2009 NEHRP RECOMMENDED SEISMIC
PROVISIONS FOR NEW BUILDINGS AND
OTHER STRUCTURES:
PART 1, PROVISIONS
Work on this 2009 edition of the NEHRP (National Earthquake Hazards Reduction Program) Recommended Seismic
Provisions for New Buildings and Other Structures began in September 2004 when the National Institute of Building
Sciences, the parent organization of the Building Seismic Safety Council (BSSC), entered into a contract with Federal
Emergency Management Agency (FEMA) for initiation of the 2009 Provisions update effort. During 2005, the BSSC member
organization representatives and alternate representatives and the BSSC Board of Direction were asked to identify
individuals to serve on the 2009 Provisions Update Committee (PUC) and its Technical Subcommittees (TSs) and to suggest
topics for concentrated study by ad hoc Issue Teams.
The 2009 PUC and its eight Technical Subcommittees (TS) then were established to address document composition and
management; design criteria and analysis and advanced technologies; mapping, foundations, and geotechnical
considerations; concrete structures; masonry structures; steel and composite steel and concrete structures; wood structures;
nonstructural components and nonbuilding structures. Three Issue Teams (ITs) also were established to focus on
performance criteria, design parameters, and foundation design requirements. Further, given ongoing work by the U.S.
Geological Survey (USGS) to update its seismic hazard maps, a Seismic Design Procedures Review Group (SDPRG) was
established to consider the emerging maps and to re-examine the existing design maps and procedures that were introduced
in the 1997 edition of the Provisions and that remained essentially unchanged for the 2000 and 2003 editions of the
Provisions.
Work already done and decisions made prior to initiation of the 2009 Provisions update project recognized that the codes
and standards arena has changed over the past decade and that those changes called for a refocusing of the Provisions on
exploration of new technologies and procedures and less consideration of format and editorial changes. To this end, the
initial efforts of the 2009 PUC and its TSs focused on adoption of the national load standard, Minimum Design Loads for
Building and Other Structures, ASCE/SEI 7-05 (including Supplements No. 1 and No. 2), as the primary reference standard
of the Provisions and the identification of parts of the 2003 Provisions that should be maintained as modifications to the
standard or otherwise revised to reflect new knowledge and experience data. The result of this effort was a vote by the BSSC
member organizations to adopt ASCE/SEI 7-05 by reference and for it to serve as the base document for the update cycle.
Three modifications to standard, originally appendices to various chapters of the 2003 Provisions, had been deemed needed
by the PUC and TSs and were approved as part of this vote by the membership for inclusion in the 2009 Provisions. As the
update cycle progressed, additional modifications to the standard were prepared and voted on by the membership in three
separate ballots. All these modifications appear in Part 1 of this document and, together with ASCE/SEI 7-05 and the
references cited therein, constitute the 2009 Provisions. (A summary of the results of the member organization ballots and
comment resolution process is available from the BSSC upon written request and will be posted on the BSSC website at
www.bssconline.org.)
A major effort also was made to rewrite the commentary to the Provisions. Until now, the commentary was published in a
separate volume and tended to explain the development of the requirements. For 2009, the commentary appears in Part 2 of
this Provisions volume and explains how to apply the Provisions requirements as articulated in ASCE/SEI 7-05 and the
references cited therein. (Note that the Part 1 modifications to the standard are accompanied by appropriate commentary
text included in Part 1.)
Part 3 of this Provisions volume introduces new procedures or provisions not currently contained in the referenced
standards for consideration and experimental use by the design community, researchers, and standards- and codedevelopment
organizations and feedback from these users is encouraged. Part 3 also presents individual summaries of
ongoing committee work that awaits additional research before being submitted to the BSSC membership for consensus
approval and provides useful guidance on the application of Part 1 requirements, either as a discussion of an overall
approach or as a detailed procedure.
1.1 INTENT
The NEHRP Recommended Seismic Provisions for New Buildings and Other Structures presents the minimum recommended
requirements necessary for the design and construction of new buildings and other structures to resist earthquake ground
motions throughout the United States. The intent of these provisions is to provide reasonable assurance of seismic
performance that will:
1. Avoid serious injury and life loss,
2. Avoid loss of function in critical facilities, and
3. Minimize structural and nonstructural repair costs where practical to do so.
These objectives are addressed by seeking to avoid structural collapse in very rare, extreme ground shaking and by seeking to
provide reasonable control of damage to structural and nonstructural systems that could lead to injury and economic or
functionality losses for more moderate and frequent ground shaking. These design requirements include minimum lateral
strength and stiffness for structural systems and guidance for anchoring, bracing, and accommodation of structural drift for
nonstructural systems.
Occupancy Category III or IV structures intended to provide enhanced safety and functionality are required to have more
strength than Occupancy Category I or II structures in an effort to reduce damage to the structural system. Allowable drifts
are reduced to control damage to building components connected to multiple floor levels. Nonstructural system performance
is enhanced by strengthening the anchorage and bracing requirements, and important equipment must be shown to be
functional after being shaken.
The degree to which these goals can be achieved depends on a number of factors including structural framing type, building
configuration, materials, as-built details, and overall quality of design. In addition, large uncertainties as to the intensity and
duration of shaking and the possibility of unfavorable response of a small subset of buildings or other structures may prevent
full realization of the intent.
1.2 REFERENCE DOCUMENT
Design for seismic resistance of structural elements including foundation elements and nonstructural components shall
conform to the requirements of ASCE/SEI 7-05, Minimum Design Loads for Buildings and Other Structures, including
Supplements No. 1 and No. 2 (referred to hereinafter as ASCE/SEI 7-05), as modified herein.1
COMMENTARY TO SECTIONS 1.1 AND 1.2
1 Supplement No. 2 of the standard is available for download at http://content.seinstitute.org/files/pdf/SupplementNo2ofthe2005Editionof
ASCE7.pdf.
2 The derivation of MCE ground motion was described in detail in Commentary Appendix A of the 2003 NEHRP Recommended
Provisions (FEMA 450-2), and this appendix, “Development of Maximum Considered Earthquake Ground Motion Maps Figures 3.3-1
through 3.3-14,” can be downloaded from http://www.nibs.org/index.php/bssc/publications/fema450nehrp2003/.
The primary intent of the NEHRP Recommended Seismic Provisions for New Buildings and Other Structures is to prevent,
for typical buildings and structures, serious injury and life loss caused by damage from earthquake ground shaking. Most
earthquake injuries and deaths are caused by structural collapse; therefore, the major thrust of the Provisions is to prevent
collapse for very rare, intense ground motion, termed the maximum considered earthquake (MCE) ground motion.2 The
intent remains the same in the 2009 Provisions; however, the prevention of collapse is redefined in terms of risk-targeted
maximum considered earthquake (MCER) ground motions. This change is explained fully in the commentary to the Part 1
modification to ASCE/SEI 7-05 Section 11.2.
Falling exterior walls and cladding and falling interior ceilings, light fixtures, pipes, equipment, and other nonstructural
components also cause deaths and injuries. The Provisions minimizes this risk using requirements for anchoring and bracing
nonstructural components, although this level of protection generally is aimed at ground motions less severe than the MCER
ground motion. This anchoring and bracing of nonstructural systems coupled with reasonable limitations on differential
movement between floors (i.e., story drift limits) also serve to control damage that may be costly to repair or that would
result in lengthy building closures, particularly for moderate shaking levels.
Stricter story drift limits can further limit damage to components connected to more than one floor (e.g., walls, cladding and
stairways) but, at the same time, can create higher acceleration levels in the building that could increase damage to
nonstructural components braced or anchored to a single floor (e.g., ceilings, light fixtures, and pipes). Achieving an
optimum balance between the cost and performance of the structural system and the effect of structural stiffness on
performance of the nonstructural systems is impossible using the prescriptive rules of a building code, particularly given the
variety of structural systems used in the United States.
Buildings deemed to have higher importance due to hazardous contents or critical occupancy are assigned to higher
Occupancy Categories (see ASCE/SEI 7-05 Table 1-1). The damage level in these buildings is intended to be reduced by
decreasing nonlinear demand using an importance factor, I, to reduce the response modification coefficient, R. The resulting
increased strength will reduce structural damage, or increase reliability of acceptable performance, for a given level of
shaking. Some authorities having jurisdiction subject the design and construction of such buildings to a higher level of
scrutiny.
The performance of critical occupancy structures in past earthquakes indicates that the increase in the importance factor
controls structural damage in moderate shaking. In strong shaking associated with the design level of two-thirds the
maximum considered earthquake or higher, the values of I have not been well tested for their effect on either functionality for
critical buildings or increased reliability of life safety protection for high occupancy buildings.
The importance factor also increases the design anchorage and bracing load for nonstructural systems, which should increase
the reliability of their staying in place and, thus, remaining undamaged. In addition, certain critical equipment must remain
operable after strong shaking. Experience data show that some nonstructural components will remain functional if they stay
in position, but other components will require testing to show that they will function following strong shaking. The emphasis
to date has been on the seismic qualification of individual components. However, the nonstructural systems of many
buildings are, in reality, complex networks that can be shut down by a single failure. For example, a break in a pressurized
pipe can flood part or all of a building forcing it to close, and failure of the anchorage (or internal workings) of a battery, day
tank, fuel lines, muffler, or main engine can shut down an emergency generator. Therefore, the special regulations for
seismic protection of nonstructural systems represent a rational approach to achieving performance appropriate for the
various occupancies, but experience data to confirm their adequacy are lacking.
When the hazard definition for design was changed from motion with a 2 percent chance of exceedance in 50 years to the 1
percent chance of collapse in 50 years, the primary intended performance was retained. The design basis ground motion is
still two-thirds of the risk-targeted maximum considered earthquake (MCER) ground motion. The increase in the importance
factor is intended to ensure a lower probability of collapse for the performance of higher occupancy and critical buildings.
The Provisions requirements are not intended to prevent damage due to earth slides (such as those that occurred in
Anchorage, Alaska) or tsunami (such as occurred in Hilo, Hawaii, and the Indian Ocean). They provide only for required
resistance to earthquake ground shaking without significant settlement, slides, subsidence, or faulting in the immediate
vicinity of the structure. In most cases, practical engineering solutions are available to resist other potential earthquake
hazards, but they must be developed on a case-by-case basis.
Although the Provisions sets the minimum performance goals described in Section 1.1, earthquake performance of buildings
and other structures is highly variable. The characteristics of the shaking itself are highly uncertain and even different sets of
motions defined to qualify as maximum considered earthquake ground motions can result in significantly different responses.
Additional uncertainty is created by the wide variety of systems and configurations allowed under the regulations as well as
by the various interpretations and implementation practices of individual designers. Thus, a small percentage of buildings
designed to the requirements of the Provisions may not meet the performance intent when exposed to earthquake ground
motions. The commentary the Tentative Provisions for the Development of Seismic Regulations for Buildings (Applied
Technology Council, 1978), upon which the first edition of the NEHRP Recommended Provisions (1985) was based,
suggested a less than 1 percent chance of collapse in a 50-year period for a building designed using the tentative
requirements. More recent studies (e.g., Quantification of Building Seismic Performance Factors, FEMA P-695, 2009)
suggest a 10 percent chance of collapse with shaking at the maximum considered earthquake level, which is roughly
equivalent to the 1978 estimations.
1.3 MODIFICATIONS TO ASCE/SEI 7-05
With only a few exceptions (such as the changes to Table 12.2-1 shown in underline and strikeout), modifications are
presented as replacements for existing sections of ASCE/SEI 7-05 or as new sections to be added to the standard.
Commentary, if any, to the modifications that appear in the remainder of this part of the 2009 Provisions is presented at
the end of each chapter of modifications. Commentary to the seismic chapters (Chapters 11 through 22) of the
unmodified reference document itself, ASCE/SEI 7-05, is presented in Part 2 of the 2009 Provisions.
Modifications to Chapter 11, Seismic Design Criteria
Replace with the following:
SECTION 11.1.2, SCOPE
11.1.2 Scope. Every structure, and portion thereof, including nonstructural components, shall be designed and
constructed to resist the effects of earthquake motions as prescribed by the seismic requirements of this standard.
Certain nonbuilding structures, as described in Chapter 15, are also within the scope and shall be designed and
constructed in accordance with the requirements for Chapter 15. Requirements concerning alterations, additions, and
change of use are set forth in Appendix 11B. Existing structures and alterations to existing structures need only comply
with the seismic requirements of this standard where required by Appendix 11B. The following structures are exempt
from the seismic requirements of this standard:
1. Detached one- and two-family dwellings that are located where the mapped, short period, spectral response
acceleration parameter, SS, is less than 0.4 or where the Seismic Design Category determined in accordance with
Section 11.6 is A, B or C.
2. Dwellings of wood-frame construction satisfying the limitations of and constructed in accordance with the
International Residential Code.
3. Buildings of wood-frame construction satisfying the limitations of and constructed in accordance with Section 2308
of the International Building Code.
4. Agricultural storage structures that are intended only for incidental human occupancy.
5. Structures that require special consideration of their response characteristics and environment that are not addressed
in Chapter 15 and for which other regulations provide seismic criteria, such as vehicular bridges, electrical
transmission towers, hydraulic structures, buried utility lines and their appurtenances, and nuclear reactors.
SECTION 11.2, DEFINITIONS
Change the definition for “Maximum Considered Earthquake (MCE) Ground Motion” to:
RISK-TARGETED MAXIMUM CONSIDERED EARTHQUAKE (MCER) GROUND MOTION: The most
severe earthquake effects considered by this standard as defined in Section 11.4.
Add the following new definition:
MAXIMUM CONSIDERED EARTHQUAKE GEOMETRIC MEAN PEAK GROUND ACCELERATION
(PGAM): The most severe earthquake effects considered for liquefaction as defined in Section 11.8.
SECTION 11.3, NOTATION
Add the following:
CR = risk coefficient; see Section 21.2.1.1
CRS = mapped value of the risk coefficient at short periods as defined by Figure 22-3
CR1 = mapped value of the risk coefficient at a period of 1 second as defined by Figure 22-4
SSD = mapped deterministic, 5 percent damped, spectral response acceleration parameter at short periods as
defined in Section 11.4.1
SSUH = mapped uniform-hazard, 5 percent damped, spectral response acceleration parameter at short periods as
defined in Section 11.4.1
S1D = mapped deterministic, 5 percent damped, spectral response acceleration parameter at a period of 1 second
as defined in Section 11.4.1
S1UH = mapped uniform-hazard, 5 percent damped, spectral response acceleration parameter at a period of 1
second as defined in Section 11.4.1
Revise the following to read as indicated:
SS = 5 percent damped, spectral response acceleration parameter at short periods as defined in Section 11.4.3
S1 = spectral response acceleration parameter at a period of 1 second as defined in Section 11.4.3
SaM = the site-specific MCER spectral response acceleration at any period
SMS = the MCER, 5 percent damped, spectral response acceleration parameter at short periods adjusted for target
risk and site-class effects as defined in Section 11.4.3
SM1 = the MCER, 5 percent damped, spectral response acceleration parameter at a period of 1 second adjusted for
target risk and site-class effects as defined in Section 11.4.3
SECTION 11.4, SEISMIC GROUND MOTION
Replace with the following:
11.4 SEISMIC GROUND MOTION VALUES
(11.4-1)
(11.4-2)
and the spectral response acceleration at a period of 1 second (S1), adjusted for the target risk of collapse, shall be
determined as the lesser value of Equations 11.4-3 and 11.4-4:
(11.4-3)
(11.4-4)
where
SSD = mapped deterministic, 5 percent damped, spectral response acceleration parameter at short periods as defined in
Section 11.4.1
SSUH = mapped uniform-hazard, 5 percent damped, spectral response acceleration parameter at short periods as defined
in Section 11.4.1
CRS = mapped value of the risk coefficient at short periods as defined in Section 11.4.1
S1D = mapped deterministic, 5 percent damped, spectral response acceleration parameter at a period of 1 second as
defined in Section 11.4.1
3
11.4.1 Mapped Acceleration Parameters and Risk Coefficients. The parameters SSUH , S1UH, SSD, and S1D shall be
determined from the 0.2- and 1-second spectral response accelerations shown on Figures 22-1 and 22-2 and Figures 22-5
through 22-6, respectively, and the risk coefficients CRS and CR1 shall be determined from Figures 22-3 and 22-4,
respectively.
11.4.2 Site Class. Based on the site soil properties, the site shall be classified as either Site Class A, B, C, D, E, or F in
accordance with Chapter 20. Where the soil properties are not known in sufficient detail to determine the Site Class, Site
Class D shall be used unless the authority having jurisdiction or geotechnical data determines Site Class E or F soils are
present at the site.
11.4.3 Site Coefficients, Risk Coefficients, and Risk-targeted Maximum Considered Earthquake (MCER) Spectral
Response Acceleration Parameters. The spectral response acceleration for short periods (SS), adjusted for the target
risk of collapse, shall be determined as the lesser value of Equations 11.4-1 and 11.4-2:
Equation
SS = CRS SSUH Equation
S SD S = S Equation
1 R1 1UH S = C S Equation
1 1D S = S
3 To utilize the U.S. Geological Survey’s seismic design map web application to obtain ground motion values, visit
http://earthquake.usgs.gov/designmaps/usapp. Also see the USGS introduction to the web application included on the CD that
accompanies this volume.
S1UH = mapped uniform-hazard, 5 percent damped, spectral response acceleration parameter at a period of 1 second as
defined in Section 11.4.1
CR1 = mapped value of the risk coefficient at a period of 1 second as defined in Section 11.4.1
The MCER spectral response acceleration for short periods (SMS) and at 1 second (SM1), adjusted for Site Class effects and
the target risk of collapse, shall be determined by Equations 11.4-5 and 11.4-6, respectively.
SMS = FaSs (11.4-5)
SM1 = FvS1 (11.4-6)
where site coefficients Fa and Fv are defined in Tables 11.4-1 and 11.4-2, respectively. When the simplified design
procedure of Section 12.14 is used, the value Fa shall be determined in accordance with Section 12.14.8.1, and the values
of Fv, S1, SMS, and SM1 need not be determined. Where S1 is less than or equal to 0.04 and SS is less than or equal to 0.15,
the structure is permitted to be assigned to Seismic Design Category A and is only required to comply with Section 11.7.
Table 11.4-1 Site Coefficient, Fa
Site
Class
Spectral Response Acceleration Parameter at Short Period
SS = 0.25
SS = 0.5
SS = 0.75
SS = 1.0
SS = 1.25
A
0.8
0.8
0.8
0.8
0.8
B
1.0
1.0
1.0
1.0
1.0
C
1.2
1.2
1.1
1.0
1.0
D
1.6
1.4
1.2
1.1
1.0
E
2.5
1.7
1.2
0.9
0.9
F
See Section 11.4.7
Note: Use straight-line interpolation for intermediate values of SS.
Table 11.4-2 Site Coefficient, Fv
Site
Class
Spectral Response Acceleration Parameter at 1-second Period
S1 = 0.1
S1 = 0.2
S1 = 0.3
S1 = 0.4
S1 = 0.5
A
0.8
0.8
0.8
0.8
0.8
B
1.0
1.0
1.0
1.0
1.0
C
1.7
1.6
1.5
1.4
1.3
D
2.4
2.0
1.8
1.6
1.5
E
3.5
3.2
2.8
2.4
2.4
F
See Section 11.4.7
Note: Use straight-line interpolation for intermediate values of S1.
11.4.4 Design Spectral Acceleration Parameters. Design earthquake spectral response acceleration parameter at short
periods, SDS, and at a 1-second period, SD1, shall be determined from Equations 11.4-7 and 11.4-8, respectively. Where
the alternate simplified design procedure of Section 12.14 is used, the value of SDS shall be determined in accordance
with Section 12.14.8.1, and the value of SD1 need not be determined:
(11.4-7)
(11.4-8)
11.4.5 Design Response Spectrum. Where a design response spectrum is required by this standard and site-specific
ground motion procedures are not used, the design response spectrum curve shall be developed as indicated in Figure
11.4-1 and as follows: Equation
2
DS 3 MS S = S Equation
1 1
2
D 3 M S = S
1. For periods less than T0, the design spectral response acceleration, Sa, shall be taken as given by Equation 11.4-9:
(11.4-9) Equation
0
0 4 0.6 a DS
S S . T
T
. .
= +..
. .
2. For periods greater than or equal to T0 and less than or equal to TS, the design spectral response acceleration, Sa, shall
be taken equal to SDS.
Figure 11.4-1 Showing design response spectrum.
1.0
0
1
0 1
Period, T (sec)
Spectral Response Acceleration, Sa (g)
SDS
SD1
D1
a
S S
T
=
T L
2
1T
S SD TL
a
·
=
T 0 T S
3. For periods greater than TS, and less than or equal to TL, the design spectral response acceleration, Sa, shall be taken
as given by Equation 11.4-10:
(11.4-10) Equation
D1
a
S S
T
=
4. For periods greater than TL, Sa shall be taken as given by Equation 11.4-11:
(11.4-11) Equation
Equation
where
SDS = the design spectral response acceleration parameter at short periods;
SD1 = the design spectral response acceleration parameter at 1-second period;
T = the fundamental period of the structure, seconds;
T0 = 0.2 SD1/SDS;
TS = SD1/SDS; and
TL = long-period transition period (seconds) shown in Figure 22-7.
11.4.6 MCER Response Spectrum. Where a MCER response spectrum is required, it shall be determined by
multiplying the design response spectrum by 1.5.
11.4.7 Site-Specific Ground Motion Procedures. The site-specific ground motion procedures set forth in Chapter 21
are permitted to be used to determine ground motions for any structure. A site response analysis shall be performed in
accordance with Section 21.1 for structures on Site Class F sites, unless the exception to Section 20.3.1 is applicable.
For seismically isolated structures and for structures with damping systems on sites with S1 greater than or equal to 0.6, a
ground motion hazard analysis shall be performed in accordance with Section 21.2.
Figure 11.4-1 Design response spectrum.
SECTION 11.8, GEOLOGIC HAZARDS AND GEOTECHNICAL INVESTIGATION
Replace with the following:
11.8 Geologic Hazards and Geotechnical Investigation
11.8.1 Site Limitation for Seismic Design Categories E and F. A structure assigned to Seismic Design Category E or
F shall not be located where there is a known potential for an active fault to cause rupture of the ground surface at the
structure.
11.8.2 Geotechnical Investigation Report Requirements for Seismic Design Categories C through F. A
geotechnical investigation report shall be provided for a structure assigned to Seismic Design Category C, D, E, or F in
accordance with this section. An investigation shall be conducted and a report shall be submitted that shall include an
evaluation of the following potential geologic and seismic hazards:
1. Slope instability;
2. Liquefaction;
3. Total and differential settlement; and
4. Surface displacement due to faulting or seismic-induced lateral spreading or lateral flow.
The report shall contain recommendations for appropriate foundation designs or other measures to mitigate the effects of
the above hazards.
EXCEPTION: Where deemed appropriate by the authority having jurisdiction, a site-specific geotechnical report is
not required when prior evaluations of nearby sites with similar soil conditions provide sufficient direction relative to
the proposed construction.
11.8.3 Additional Geotechnical Investigation Report Requirements for Seismic Design Categories D through F.
The geotechnical investigation report for a structure assigned to Seismic Design Category D, E, or F shall include:
1. The determination of dynamic seismic lateral earth pressures on basement and retaining walls due to design
earthquake ground motions.
2. The potential for liquefaction and soil strength loss evaluated for site peak ground acceleration, earthquake
magnitude, and source characteristics consistent with the maximum considered earthquake geometric mean peak
ground accelerations. Peak ground acceleration shall be determined based on either: (a) a site-specific study taking
into account soil amplification effects as specified in Section 11.4.7 or (b) the peak ground acceleration, PGAM, from
Equation 11.8-1:
PGAM = FPGA PGA (11.8-1)
where
PGAM = maximum considered earthquake geometric mean peak ground acceleration adjusted for Site Class
effects;
PGA = mapped maximum considered earthquake geometric mean peak ground acceleration shown in Figures
22-8 through 22-11; and
FPGA = site coefficient from Table 11.8-1.
3. Assessment of potential consequences of liquefaction and soil strength loss as computed in Item 2, including
estimation of total and differential settlement, lateral soil movement, lateral soil loads on foundations, reduction in
foundation soil-bearing capacity and lateral soil reaction, soil downdrag and reduction in axial and lateral soil
reaction for pile foundations, increases in soil lateral pressures on retaining walls, and flotation of buried structures.
4. Discussion of mitigation measures such as selection of appropriate foundation type and depths, selection of
appropriate structural systems to accommodate anticipated displacements and forces, ground stabilization, or any
combination of these measures and how they shall be considered in the design of the structure.
Table 11.8-1 Site Coefficient FPGA
Site
Class
Mapped MCE Geometric Mean Peak Ground Acceleration, PGA
PGA = 0.1
PGA = 0.2
PGA = 0.3
PGA = 0.4
PGA = 0.5
A
0.8
0.8
0.8
0.8
0.8
B
1.0
1.0
1.0
1.0
1.0
C
1.2
1.2
1.1
1.0
1.0
D
1.6
1.4
1.2
1.1
1.0
E
2.5
1.7
1.2
0.9
0.9
F
See Section 11.4.7
Note: Use straight-line interpolation for intermediate values of PGA.
Commentary to Chapter 11 Modifications
COMMENTARY TO SECTION 11.1.2
C11.1.2 Scope. The scope statement establishes in general terms the applicability of the standard as a base of reference.
Certain structures are exempt and need not comply. The reasons for each are described below.
Note that it is not acceptable to use a combination of International Building Code (IBC) and International Residential Code
(IRC) conventional construction provisions. Conventional requirements of either the IBC or the IRC can be combined with
engineered design of elements in accordance with IBC engineered design requirements. Elements designed using the IBC
engineered design requirements are not exempt from the seismic requirements of ASCE/SEI 7.
Exemption 1 – Detached one- and two-family dwellings in Seismic Design Categories A, B, and C, along with those located
where Ss < 0.4g, are exempt because they represent low seismic risks.
Exemption 2 – This exemption recognizes that the wood-frame seismic design requirements of the International Residential
Code (IRC) substantially meet the intent of conventional construction (wood-frame) provisions included in the NEHRP
Recommended Seismic Provisions through the 2003 Edition.
Exemption 3 – This exemption recognizes that wood-frame seismic design requirements of International Building Code
(IBC) Section 2308 substantially meet the intent of conventional construction (wood-frame) provisions included in the
NEHRP Recommended Seismic Provisions through the 2003 Edition.
Exemption 4 – Agricultural storage structures generally are exempt from most code requirements because of the
exceptionally low risk to life involved.
Exemption 5 – Bridges, transmission towers, nuclear reactors, and other structures with special configurations and uses are
not covered because regulations developed to apply to buildings and building-like structures do not adequately address their
design and performance issues.
The standard is not retroactive and usually applies to existing structures only where there is an addition, change of use, or
alteration. Minimum acceptable seismic resistance of existing buildings is a policy issue normally set by the authority having
jurisdiction. Appendix 11B of the standard contains rules of application for basic conditions. ASCE 31, Seismic Evaluation
of Buildings, and ASCE 41, Seismic Rehabilitation of Existing Buildings, are available for technical guidance but do not
contain policy recommendations. The International Code Council includes a chapter in the IBC to control the alteration,
repair, addition, and change of occupancy of existing buildings and also maintains the International Existing Building Code
(IEBC) and an associated commentary.
COMMENTARY TO SECTION 11.2
C11.2 DEFINITIONS
Renaming the maximum considered earthquake (MCE) ground motions as the risk-targeted maximum considered earthquake
(MCER) ground motions is an editorial change recommended by the BSSC’s Provisions Update Committee and accepted by
the BSSC’s Board. The MCER ground motions are based on the 2008 USGS seismic hazard maps and also incorporate three
technical changes to ASCE/SEI 7-05:
1. Use of risk-targeted ground motions,
2. Use of maximum direction ground motions, and
3. Use of near-source 84th percentile ground motions.
Reasons for each of the three technical changes are included in the commentary that accompanies the modifications to
Chapter 21.
COMMENTARY TO SECTIONS 11.4.3 AND 11.4.4
C11.4.3 Site Coefficients, Risk Coefficients, and Risk-targeted Maximum Considered Earthquake (MCER) Spectral
Response Acceleration Parameters. The following illustrates the process of developing MCER response spectral
accelerations using the formulas and maps of Section 11.4.3 and Chapter 22, respectively, and provides a summary of design
ground motions for 34 city sites in regions of the United States of greatest seismic risk. Additional information and
references explaining the differences from the MCE ground motions in ASCE/SEI 7-05 are included in the commentary to
Chapter 21.
Illustration of the Development of MCER Spectral Response Acceleration Using Section 11.4.3 Equations and Chapter 22
Uniform-Hazard, Risk Coefficient, and Deterministic Maps.
The formulas of Section 11.4.3 (and the associated uniform-hazard, risk coefficient, and deterministic maps of Chapter 22)
are intended to add transparency to the development of MCER ground motions. The development of MCER ground motions
is explained in Section 21.2 and its commentary as part of the site-specific ground motion procedures for seismic design. As
will be illustrated, the formulas (and maps) add transparency by emulating the site-specific procedure. A cost of this
transparency is the added complexity of more formulas (and maps). However, a USGS website similar to the USGS Java
ground motion parameter calculator automates use of the proposed formulas (and maps):
http://earthquake.usgs.gov/designmaps/usapp.
The three steps that the website implements are as follows:
Step 1 – Adjust uniform-hazard ground motions (Site Class B) for target risk of collapse
As illustrated in the top row of Figure C11.4-1, the first step is to obtain the mapped uniform-hazard (2 percent-in-50-years
probability of exceedance) spectral response acceleration for short periods (SSUH) from Figure 22-1 and for a period of 1
second (S1UH) from Figure 22-2, and then to multiply these values by the corresponding mapped risk coefficients (CRS and
CR1) from Figures 22-3 and 22-4, respectively. This step is expressed in Equations 11.4-1 for the short periods and 11.4-3 for
the 1-second period and is consistent with Section 21.2.1 of the site-specific procedure in Chapter 21. The resulting spectral
response accelerations, CRS, SSUH and CR1S1UH, are referred to as probabilistic ground motions. Figure C11.4-1 illustrates this
for the 1-second period only using small maps of the conterminous United States that depict S1UH, CR1, and CR1S1UH.
The reasons for using 2 percent-in-50-years (uniform-hazard) spectral response accelerations, which were the basis for the
probabilistic portions of the MCE ground motion maps in ASCE/SEI 7-05, are explained in the commentary of the 2003
NEHRP Recommended Provisions. As explained below in the Chapter 21 commentary, the uniform-hazard maps (Figures
22-1 and 22-2) represent the spectral response acceleration in the maximum direction, which are larger than the geometric
mean spectral response acceleration maps developed by the USGS by factors of 1.1 for the short periods and 1.3 for the 1-
second period. The risk coefficients adjust these uniform-hazard (2 percent-in-50-years) spectral response accelerations to
achieve building designs with 1 percent probability of collapse in 50 years (i.e., uniform risk), as explained below in the
Chapter 21 commentary.
Step 2 – Take minimum of probabilistic and deterministic ground motions (Site Class B)
As illustrated in the middle row of Figure C11.4-1, the second step in the development of MCER ground motions is to obtain
the mapped deterministic spectral response acceleration for short periods (SSD) from Figure 22-5 and for a period of 1 second
(S1D) from Figure 22-6, and then to take the minimum of each of these values (expressed in Equations 11.4-2 and 11.4-4,
respectively) and the corresponding value resulting from Step 1 (i.e., those expressed in Equations 11.4-1 and 11.4-3,
respectively). This step is consistent with Sections 21.2.2 (“Deterministic Ground Motions”) and 21.2.3 (“Site Specific
MCER”) of the site-specific procedure in Chapter 21. The resulting spectral response accelerations are denoted SS for the
short periods and S1 for the 1-second period. Figure C11.4-1 illustrates this for the 1-second period only using small maps of
the conterminous United States that depict CR1S1UH, S1D, and S1.
The reasons for using the minimum of probabilistic and deterministic spectral response accelerations, which was done
previously (but not transparently) in developing the MCE ground motions maps in ASCE/SEI 7-05, are explained in the
commentary of the 2003 NEHRP Recommended Provisions. In brief, deterministic ground motions provide a reasonable and
practical upper-bound to design ground motions, but their use implies a somewhat higher level of collapse risk than the 1
percent probability of collapse in 50 years associated with probabilistic (risk-targeted) ground motions. In general,
deterministic ground motions govern only at sites near active sources in regions of high seismicity.
As defined in ASCE/SEI 7-05 Section 21.2.2, the deterministic spectral response accelerations (for Site Class B) shall not be
taken as lower than 1.5g for the short periods and 0.6g for the 1-second period; hence, the ground motions on the proposed
deterministic maps (Figures 22-5 and 22-6) are no lower than these values. Otherwise the ground motions on the proposed
deterministic maps are 180 percent (as opposed to 150 percent in ASCE/SEI 7-05) of median spectral response accelerations,
for reasons explained below in the section entitled “Deterministic Ground Motions – 84th Percentile.” Like the proposed
uniform-hazard maps used in Step 1, the proposed deterministic maps represent the spectral response acceleration in the
maximum direction.
Step 3 – Adjust Site Class B ground motions for site condition (e.g., Site Class D)
As illustrated in the bottom row of Figure C11.4-1, the third step is to multiply the spectral response accelerations resulting
from Step 2 (SS and S1) by the corresponding site coefficients (Fa and Fv) from Tables 11.4-1 and 11.4-2, respectively. This
step is expressed in Equation 11.4-5 for the short periods and 11.4-6 for the 1-second period, where the resulting ground
motions are named risk-targeted maximum considered earthquake (MCER) spectral response accelerations and are denoted
SMS and SM1, respectively. Figure C11.4-1 illustrates the step for the 1-second period only using a small map of the
conterminous United States that depicts S1, an abbreviated version of Table 11.4-2, and another small map that depicts SM1.
This step is the same as that in ASCE/SEI 7-05 Section 11.4.3, except that the resulting MCE spectral response accelerations
(SMS and SM1) have been renamed MCER spectral response accelerations.
Figures C11.4-2 and C11.4-3 are maps of the United States and California, respectively, showing values of the MCER 1-
second spectral response acceleration parameter, SM1, and associated regions of Seismic Design Category, assuming Site
Class D conditions. These maps illustrate MCER ground motions resulting from the three-step process described above for
the 1-second period only.
The design ground motions are 2/3 of these MCER ground motions as calculated using Equations 11.4-7 and 11.4-8.
Summary of Design Ground Motions – 34 United States Cities
Example values of the design ground motions that incorporate both USGS updates to uniform-hazard values (and hazard
functions), including the new NGA relations, and the three technical changes mentioned above, are shown next. For
comparison, values of design ground motions of the current standard (ASCE/SEI 7-05) and, for California sites, values of
design ground motions of the 2001 California Building Code (CBC) are given. In all cases, example values are based on
design ground motions, representative of Site Class D conditions (i.e., default site class).
Table C11.4-1 lists the 34 city sites by region, the county (or counties) and associated populations they represent, and the
latitude and longitude of the specific location of the city site. Typically, each city is the largest city of the county or
metropolitan statistical area (MSA) of interest. The exception is Los Angeles County which has four city sites due to its large
geographical area and associated risk. The specific location (latitude and longitude) of city sites is important for sites in high
seismic regions (i.e., near an active source) since ground motions can vary greatly over relatively small distances. Example
sites are selected to be coincident with the location of the hazard grid point nearest the center of the city of interest. Hazard
grid points are the discrete locations at which the USGS calculates values of probabilistic and deterministic ground motions
(and risk coefficients). At the time that these examples were developed, ground motions were available (from the USGS)
only for these discrete locations; however, final maps and database tools such as the USGS online ground motion parameter
calculator also provide values of ground motions at intermediate locations.
Table C11.4-2 provides values of short-period spectral acceleration, SDS, and Table C11.4-3 provides values of 1-second
spectral acceleration, SD1, for each of the 34 city sites of Table C11.4-1. Spectral acceleration values and Seismic Design
Category (SDC) are given for both ASCE/SEI 7-05 provisions and changes put forth in these provisions (2009 Provisions).
For California city sites, these tables also provide the corresponding values of seismic coefficients (2.5Ca, at short periods,
and, Cv, at 1 second) of the 2001 California Building Code (1997 Uniform Building Code or UBC). Weighted mean values
of spectral acceleration (and seismic coefficients) are calculated for each region considering the population associated with
each city site in Tables 11.4-2 and 11.4-3.
The following observations are made by comparing the design ground motions of these Provisions with those of ASCE/SEI
7-05 and the design coefficients of the 2001 California Building Code (CBC):
1. On a regional basis, the changes to ASCE/SEI 7-05 put forth in these Provisions result in only a slight increase or
decrease in design ground motions, on average. Notable exceptions are short-period ground motions in the central and
eastern United States (CEUS) for which the changes reduce design values and for certain city sites (e.g., St. Louis,
Chicago, and New York) where the changes also lower the Seismic Design Category.
2. In the western region (WUS), the changes to ASCE/SEI 7-05 put forth in these Provisions result in a modest increase, or
decrease, in design ground motions (plus or minus 10 percent), and generally lower seismic design values from those of
2001 CBC (1997 UBC).
3. For certain city sites (e.g., San Bernardino and San Diego), the changes to ASCE/SEI 7-05 put forth in these Provisions
result in a substantial increase, or decrease, in design ground motions due primarily to changes in underlying updated
USGS hazard functions.
Step 1 Figure C11.4-1 Illustration of process for developing 1-second MCER Site Class D ground motions using formulas of Section 11.4.3 and associated mapped values of ground motions and risk coefficients of Chapter 22.
Step 2 Figure C11.4-1 Illustration of process for developing 1-second MCER Site Class D ground motions using formulas of Section 11.4.3 and associated mapped values of ground motions and risk coefficients of Chapter 22.
Step 3 Figure C11.4-1 Illustration of process for developing 1-second MCER Site Class D ground motions using formulas of Section 11.4.3 and associated mapped values of ground motions and risk coefficients of Chapter 22.
Figure C11.4-1 Illustration of process for developing 1-second MCER Site Class D ground motions using formulas of
Section 11.4.3 and associated mapped values of ground motions and risk coefficients of Chapter 22.
S1 – Map of Site Class B
ground motions
Fv – Site coefficient
(Table 11.4-2)
SM1 – Map of MCER ground
motions (Site Class D)
Figure C11.4-2 Map illustrating values of the MCER 1-second spectral response acceleration parameter, SM1 (%g), and associated regions of Seismic Design Category, assuming Site Class D conditions.
Figure C11.4-2 Map illustrating values of the MCER 1-second spectral response acceleration parameter, SM1
(%g), and associated regions of Seismic Design Category, assuming Site Class D conditions.
Figure C11.4-3 Map illustrating values of the MCER 1-second spectral response acceleration parameter, SM1 (%g), and associated regions of Seismic Design Category, assuming Site Class D conditions, for California sites.
Figure C11.4-3 Map illustrating values of the MCER 1-second spectral response acceleration parameter, SM1
(%g), and associated regions of Seismic Design Category, assuming Site Class D conditions, for California
sites.
Table C11.4-1 Showing Thirty-Four Cities, Site Locations (Latitude and Longitude), and Associated Counties and Populations At Risk for Which Values of Ground Motions Are Provided
Table C11.4-1 Thirty-Four Cities, Site Locations (Latitude and Longitude), and Associated Counties and Populations At Risk for
Which Values of Ground Motions Are Provided
Table C11.4-2 Showing Comparison of Values of the Short-Period Design Ground Motion Parameter (SDS) and Corresponding Seismic Design Category (SDC) Put Forth in These Provisions with ASCE/SEI 7-05 and 1997 UBC Values for 34 City Site Locations (Assuming Default Site Class D)
Table C11.4-2 Comparison of Values of the Short-Period Design Ground Motion Parameter (SDS) and Corresponding Seismic Design
Category (SDC) Put Forth in These Provisions with ASCE/SEI 7-05 and 1997 UBC Values for 34 City Site Locations (Assuming
Default Site Class D)
Table C11.4-3 Showing Comparison of Values of the 1-Second Period Design Ground Motion Parameter (SD1) and Corresponding Seismic Design Category (SDC) Put Forth in These Provisions with ASCE/SEI 7-05 and 1997 UBC Values for 34 City Site Locations (Assuming Default Site Class D)
Table C11.4-3 Comparison of Values of the 1-Second Period Design Ground Motion Parameter (SD1) and Corresponding Seismic
Design Category (SDC) Put Forth in These Provisions with ASCE/SEI 7-05 and 1997 UBC Values for 34 City Site Locations
(Assuming Default Site Class D)
COMMENTARY TO SECTION 11.8.3
C11.8.3 Additional Geotechnical Investigation Report Requirements for Seismic Design Categories D through F. The
dynamic lateral earth pressure on basement and retaining walls during the period of earthquake ground shaking is considered
to be an earthquake load, E, for use in design load combinations. This dynamic earth pressure is superimposed on the preexisting
static lateral earth pressure during ground shaking. The pre-existing static lateral earth pressure is considered to be
an H load.
While the dynamic seismic lateral earth pressures (Item 1) may be determined for design earthquake ground motions, taken
as 2/3 of maximum considered earthquake ground motions, the potential for liquefaction and soil strength loss and related
consequences (Items 2 and 3) must be evaluated for maximum considered earthquake ground motions because they can be
catastrophic to a structure.
Page intentionally left blank.
Modifications to Chapter 12, Seismic Design Requirements
for Building Structures
TABLE 12.2-1, DESIGN COEFFICIENTS AND FACTORS
FOR SEISMIC-FORCE-RESISTING SYSTEMS
Revise as indicated (substantive changes are shaded, deletions are shown in strikeout, and
additions are underlined):
Table 12.2-1 Design Coefficients and Factors for Seismic-Force-Resisting Systems
Seismic-Force-Resisting
System
ASCE/SEI 7-05
Section Where
Detailing
Requirements
Are Specified
Response
Modification
Coefficient,
Ra
System
Overstren
gth Factor,
O0
g
Deflection
Amplificat
ion
Factor,
Cd
b
Structural System
Limitations and Building
Height (ft) Limitc
Seismic Design Category
B
C
Dd
Ed
Fe
A. BEARING WALL SYSTEMS
1. Special reinforced concrete
shear walls
14.2 and 14.2.3.6
5
2½
5
NL
NL
160
160
100
2. Ordinary reinforced
concrete shear walls
14.2 and 14.2.3.4
4
2½
4
NL
NL
NP
NP
NP
3. Detailed plain concrete
shear walls
14.2 and 14.2.3.2
2
2½
2
NL
NP
NP
NP
NP
4. Ordinary plain concrete
shear walls
14.2 and 14.2.3.1
1½
2½
1½
NL
NP
NP
NP
NP
5. Intermediate precast shear
walls
14.2 and 14.2.3.5
4
2½
4
NL
NL
40k
40k
40k
6. Ordinary precast shear
walls
14.2 and 14.2.3.3
3
2½
3
NL
NP
NP
NP
NP
7. Special reinforced
masonry shear walls
14.4 and 14.4.3
5
2½
3½
NL
NL
160
160
100
8. Intermediate reinforced
masonry shear walls
14.4 and 14.4.3
3½
2½
2¼
NL
NL
NP
NP
NP
9. Ordinary reinforced
masonry shear walls
14.4
2
2½
1¾
NL
160
NP
NP
NP
10. Detailed plain masonry
shear walls
14.4
2
2½
1¾
NL
NP
NP
NP
NP
11. Ordinary plain masonry
shear walls
14.4
1½
2½
1¼
NL
NP
NP
NP
NP
12. Prestressed masonry
shear walls
14.4
1½
2½
1¾
NL
NP
NP
NP
NP
13. Light-framed walls
sheathed with wood
structural panels rated for
shear resistance or steel
sheets
14.1, 14.1.4.2,
and 14.5
6½
3
4
NL
NL
65
65
65
14. Light-framed walls with
shear panels of all other
materials
14.1, 14.1.4.2,
and 14.5
2
2½
2
NL
NL
35
NP
NP
15. Light-framed wall
systems using flat strap
bracing
14.1, 14.1.4.2,
and 14.5
4
2
3½
NL
NL
65
65
65
16. Ordinary reinforced AAC
masonry shear walls
14.4.5.4
2
2 ½
2
NL
35
NP
NP
NP
17. Plain AAC masonry shear
walls
14.4.5.3
1 ½
2 ½
1 ½
NL
NP
NP
NP
NP
B. BUILDING FRAME SYSTEMS
1. Steel eccentrically braced
frames, moment resisting
connections at columns
away from links
14.1
8
2
4
NL
NL
160
160
100
2. Steel eccentrically braced
frames, non-momentresisting,
connections at
columns away from links
14.1
7
2
4
NL
NL
160
160
100
23. Special steel
concentrically braced
frames
14.1
6
2
5
NL
NL
160
160
100
34. Ordinary steel
concentrically braced
frames
14.1
3¼
2
3¼
NL
NL
35j
35j
NPj
45. Special reinforced
concrete shear walls
14.2 and 14.2.3.6
6
2½
5
NL
NL
160
160
100
56. Ordinary reinforced
concrete shear walls
14.2 and 14.2.3.4
5
2½
4½
NL
NL
NP
NP
NP
67. Detailed plain concrete
shear walls
14.2 and 14.2.3.2
2
2½
2
NL
NP
NP
NP
NP
78. Ordinary plain concrete
shear walls
14.2 and 14.2.3.1
1½
2½
1½
NL
NP
NP
NP
NP
89. Intermediate precast
shear walls
14.2 and 14.2.3.5
5
2½
4½
NL
NL
40k
40k
40k
910. Ordinary precast shear
walls
14.2 and 14.2.3.3
4
2½
4
NL
NP
NP
NP
NP
101. Composite steel and
concrete eccentrically
braced frames
14.3
8
2
4
NL
NL
160
160
100
112. Composite steel and
concrete concentrically
braced frames
14.3
5
2
4½
NL
NL
160
160
100
123. Ordinary composite
steel and concrete braced
frames
14.3
3
2
3
NL
NL
NP
NP
NP
134. Composite steel plate
shear walls
14.3
6½
2½
5½
NL
NL
160
160
100
145. Special composite
reinforced concrete shear
walls with steel elements
14.3
6
2½
5
NL
NL
160
160
100
156. Ordinary composite
reinforced concrete shear
walls with steel elements
14.3
5
2½
4½
NL
NL
NP
NP
NP
167. Special reinforced
masonry shear walls
14.4
5½
2½
4
NL
NL
160
160
100
178. Intermediate reinforced
masonry shear walls
14.4
4
2½
4
NL
NL
NP
NP
NP
189. Ordinary reinforced
masonry shear walls
14.4
2
2½
2
NL
160
NP
NP
NP
1920. Detailed plain
masonry shear walls
14.4
2
2½
2
NL
NP
NP
NP
NP
201. Ordinary plain masonry
shear walls
14.4
1½
2½
1¼
NL
NP
NP
NP
NP
212. Prestressed masonry
shear walls
14.4
1½
2½
1¾
NL
NP
NP
NP
NP
223. Light-framed walls
sheathed with wood
structural panels rated for
shear resistance or steel
sheets
14.1, 14.1.4.2, and
14.5
7
2½
4½
NL
NL
65
65
65
234. Light-framed walls
with shear panels of all
14.1, 14.1.4.2, and
14.5
2½
2½
2½
NL
NL
35
NP
NP
other materials
25. Buckling-restrained
braced frames, nonmoment-
resisting beamcolumn
connections
14.1
7
2
5½
NL
NL
160
160
100
246. Buckling-restrained
braced frames, momentresisting
beam-column
connections
14.1
8
2½
5
NL
NL
160
160
100
257. Special steel plate shear
wall
14.1
7
2
6
NL
NL
160
160
100
C. MOMENT-RESISTING FRAME SYSTEMS
1. Special steel moment
frames
14.1 and 12.2.5.5
8
3
5½
NL
NL
NL
NL
NL
2. Special steel truss
moment frames
14.1
7
3
5½
NL
NL
160
100
NP
3. Intermediate steel
moment frames
12.2.5.6, 12.2.5.7,
12.2.5.8, 12.2.5.9,
and 14.1
4.5
3
4
NL
NL
35h,i
NPh
NPi
4. Ordinary steel moment
frames
12.2.5.6, 12.2.5.7,
12.2.5.8, and 14.1
3.5
3
3
NL
NL
NPh
NPh
NPi
5. Special reinforced
concrete moment frames
12.2.5.5 and 14.2
8
3
5½
NL
NL
NL
NL
NL
6. Intermediate reinforced
concrete moment frames
14.2
5
3
4½
NL
NL
NP
NP
NP
7. Ordinary reinforced
concrete moment frames
14.2
3
3
2½
NL
NP
NP
NP
NP
8. Special composite steel
and concrete moment
frames
12.2.5.5 and 14.3
8
3
5½
NL
NL
NL
NL
NL
9. Intermediate composite
moment frames
14.3
5
3
4½
NL
NL
NP
NP
NP
10. Composite partially
restrained moment
frames
14.3
6
3
5½
160
160
100
NP
NP
11. Ordinary composite
moment frames
14.3
3
3
2½
NL
NP
NP
NP
NP
12. Cold-formed steel –
special bolted framem
14.1
3½
3 l
3½
35
35
35
35
35
D. DUAL SYSTEMS
WITH SPECIAL
MOMENT FRAMES
CAPABLE OF
RESISTING AT
LEAST 25% OF
PRESCRIBED
SEISMIC FORCES
12.2.5.1
1. Steel eccentrically braced
frames
14.1
8
2½
4
NL
NL
NL
NL
NL
2. Special steel
concentrically braced
frames
14.1
7
2½
5½
NL
NL
NL
NL
NL
3. Special reinforced
concrete shear walls
14.2
7
2½
5½
NL
NL
NL
NL
NL
4. Ordinary reinforced
concrete shear walls
14.2
6
2½
5
NL
NL
NP
NP
NP
5. Composite steel and
concrete eccentrically
braced frames
14.3
8
2½
4
NL
NL
NL
NL
NL
6. Composite steel and
concrete concentrically
braced frames
14.3
6
2½
5
NL
NL
NL
NL
NL
7. Composite steel plate
14.3
7½
2½
6
NL
NL
NL
NL
NL
shear walls
8. Special composite
reinforced concrete shear
walls with steel elements
14.3
7
2½
6
NL
NL
NL
NL
NL
9. Ordinary composite
reinforced concrete shear
walls with steel elements
14.3
6
2½
5
NL
NL
NP
NP
NP
10. Special reinforced
masonry shear walls
14.4
5½
3
5
NL
NL
NL
NL
NL
11. Intermediate reinforced
masonry shear walls
14.4
4
3
3½
NL
NL
NP
NP
NP
12. Buckling-restrained
braced frame
14.1
8
2½
5
NL
NL
NL
NL
NL
13. Special steel plate shear
walls
14.1
8
2½
6½
NL
NL
NL
NL
NL
E. DUAL SYSTEMS
WITH INTERMEDIATE
MOMENT FRAMES
CAPABLE OF RESISTING
AT LEAST 25% OF
PRESCRIBED SEISMIC
FORCES
12.2.5.1
1. Special steel
concentrically braced
framesf
14.1
6
2½
5
NL
NL
35
NP
NPh,k
2. Special reinforced
concrete shear walls
14.2
6½
2½
5
NL
NL
160
100
100
3. Ordinary reinforced
masonry shear walls
14.4
3
3
2½
NL
160
NP
NP
NP
4. Intermediate reinforced
masonry shear walls
14.4
3½
3
3
NL
NL
NP
NP
NP
5. Composite steel and
concrete concentrically
braced frames
14.3
5½
2½
4½
NL
NL
160
100
NP
6. Ordinary composite
braced frames
14.3
3½
2½
3
NL
NL
NP
NP
NP
7. Ordinary composite
reinforced concrete shear
walls with steel elements
14.3
5
3
4½
NL
NL
NP
NP
NP
8. Ordinary reinforced
concrete shear walls
14.2
5½
2½
4½
NL
NL
NP
NP
NP
F. SHEAR WALLFRAME
INTERACTIVE
SYSTEM WITH
ORDINARY
REINFORCED
CONCRETE MOMENT
FRAMES AND
ORDINARY
REINFORCED
CONCRETE SHEAR
WALLS
12.2.5.10 and 14.2
4½
2½
4
NL
NP
NP
NP
NP
G. CANTILEVERED
COLUMN SYSTEMS
DETAILED TO
CONFORM TO THE
REQUIREMENTS FOR:
12.2.5.2
1. Special steel moment
frames
12.2.5.5 and 14.1
2½
1¼
2½
35
35
35
35
35
2. Intermediate steel
moment frames
14.1
1½
1¼
1½
35
35
35h
NPh,i
NPh,i
3. Ordinary steel moment
frames
14.1
1¼
1¼
1¼
35
35
NP
NPh,i
NPh,i
4. Special reinforced
concrete moment frames
12.2.5.5 and 14.2
2½
1¼
2½
35
35
35
35
35
5. Intermediate concrete
moment frames
14.2
1½
1¼
1½
35
35
NP
NP
NP
6. Ordinary concrete
moment frames
14.2
1
1¼
1
35
NP
NP
NP
NP
7. Timber frames
14.5
1½
1½
1½
35
35
35
NP
NP
H. STEEL SYSTEMS
NOT SPECIFICALLY
DETAILED FOR SEISMIC
RESISTANCE,
EXCLUDING
CANTILEVER COLUMN
SYSTEMS
14.1
3
3
3
NL
NL
NP
NP
NP
aResponse modification coefficient, R, for use throughout the standard. Note R reduces forces to a strength level, not an allowable stress
level.
bReflection amplification factor, Cd, for use in Sections 12.8.6, 12.8.7, and 12.9.2
cNL = Not Limited and NP = Not Permitted. For metric units use 30.5 m for 100 ft and use 48.8 m for 160 ft. Heights are measured
from the base of the structure as defined in Section 11.2.
dSee Section 12.2.5.4 for a description of building systems limited to buildings with a height of 240 ft (73.2 m) or less.
eSee Section 12.2.5.4 for building systems limited to buildings with a height of 160 ft (48.8 m) or less.
fOrdinary moment frame is permitted to be used in lieu of intermediate moment frame for Seismic Design Categories B or C.
gThe tabulated value of the overstrength factor, O0, is permitted to be reduced by subtracting one-half for structures with flexible
diaphragms, but shall not be taken as less than 2.0 for any structure.
hSee Sections 12.2.5.6 and 12.2.5.7 for limitations for steel OMFs and IMFs in structures assigned to Seismic Design Category D or E.
iSee Sections 12.2.5.8 and 12.2.5.9 for limitations for steel OMFs and IMFs in structures assigned to Seismic Design Category F.
jSteel ordinary concentrically braced frames are permitted in single-story buildings up to a height of 60 ft (18.3 m) where the dead load
of the roof does not exceed 20 psf (0.96 kN/m2) and in penthouse structures.
l Alternatively, the seismic load effect with overstrength, Em, can be based on the expected strength determined in accordance with AISI
S110.
m Cold-formed steel – special bolted moment frames shall be limited to one story in height in accordance with AISI S110.
TABLE 12.6-1, PERMITTED ANALYTICAL PROCEDURES
Replace with the following:
Table 12.6-1 Permitted Analytical Procedures
Seismic
Design
Category
Structural Characteristics
Equivalent
Lateral Force
Analysis Section
12.8
Modal Response
Spectrum
Analysis Section
12.9
Seismic
Response
History
Procedures
Chapter 16
B, C
All structures
P
P
P
D, E, F
Regular structures not exceeding 160 feet in
height and all structures of light frame
construction
P
P
P
Regular structures equal to or exceeding
160 feet in height with T < 3.5 Ts
P
P
P
Irregular structures not exceeding 160 feet in
height and having only horizontal
irregularities type 2, 3, 4, or 5 of Table 12.3-
1 or vertical irregularities type 4, 5a or 5b of
Table 12.3-2
P
P
P
All other structures
NP
P
P
Note: P – Permitted; NP – Not permitted.
SECTION 12.8.7, P-DELTA LIMIT
Replace with the following:
12.8.7 P-delta Limit. Stability coefficient, ., as determined for each level of the structure by the following equation,
shall not exceed 0.10:
(12.8-16)
where:
Px = the total vertical design load at and above Level x. Where calculating the vertical design load
for purposes of determining P-delta effects, the individual load factors need not exceed 1.0.
. = the design story drift calculated in accordance with Section 12.8.6.
I = the occupancy importance factor determined in accordance with Section 11.5.1.
Vx = the seismic shear force acting between Level x and x - 1.
hsx = the story height below Level x.
Cd = the deflection amplification factor from Table 12.2-1
EXCEPTION: The stability coefficient, ., shall be permitted to exceed 0.10 if either of the following applies: Equation
Equation
1. The resistance to lateral forces is determined to increase continuously in a monotonic nonlinear static
(pushover) analysis according to ASCE/SEI 41 Section 3.3.3.3.2 using Sa defined as a MCER spectral response
acceleration according to the Provisions at the effective fundamental period. Modeling and analysis shall
conform to ASCE/SEI 41 Section 3.3.3, except that the analysis shall be done for seismic actions occurring
simultaneously with the effects of dead load in combination with not less than 25 percent of the required design
live loads, reduced as permitted for the area of a single floor. Degradation shall be modeled and P-delta effects
shall be included in the analysis. A review of the nonlinear static analysis shall be performed by an independent
team having experience in seismic analysis methods and the theory and application of nonlinear seismic
analysis and structural behavior under earthquake loading. The review team shall be composed of at least two
members including at least one registered design professional.
2. Compliance with the provisions of the nonlinear response history procedure in Chapter 16 is demonstrated.
SECTION 12.11.2.2.1, TRANSFER OF ANCHORAGE FORCES INTO DIAPHRAGM
Replace with the following:
12.11.2.2.1 Transfer of Anchorage Forces into Diaphragm. Diaphragms shall be provided with continuous ties or
struts between diaphragm chords to distribute these anchorage forces into the diaphragm.
EXCEPTION: In buildings with diaphragms of light-frame construction, continuous cross-ties are not required
provided all of the following are satisfied:
1. The unsupported height of the wall does not exceed 12 feet,
2. Anchorages are spaced no more than 4 feet on center,
3. The length of the diaphragm in the direction parallel to the wall being anchored does not exceed 2.5 times the
length of the diaphragm in the orthogonal direction, and
4. The anchorage connection extends far enough into the diaphragm to transfer the anchorage force into the
diaphragm.
Diaphragm connections shall be positive, mechanical, or welded. Added chords are permitted to be used to form
subdiaphragms to transmit the anchorage forces to the main continuous cross-ties. The maximum length-to-width ratio
of the structural subdiaphragm shall be 2.5 to 1. Connections and anchorages capable of resisting the prescribed forces
shall be provided between the diaphragm and the attached components. Connections shall extend into the diaphragm a
sufficient distance to develop the force transferred into the diaphragm.
SECTION 12.11.2.2.3, WOOD DIAPHRAGMS
Replace with the following:
12.11.2.2.3 Wood Diaphragms. In wood diaphragms, the continuous ties shall be in addition to the diaphragm
sheathing.
EXCEPTION: Where continuous cross-ties are not required by Section 12.11.2.2.1 and the anchorage connections
extend into the diaphragm a sufficient distance to develop the force transferred into the diaphragm sheathing.
Anchorage shall not be accomplished by use of toenails or nails subject to withdrawal nor shall wood ledgers or
framing be used in cross-gain bending or cross-grain tension. The diaphragm sheathing shall not be considered
effective as providing the ties or struts required by this section.
SECTION 12.14.7.5.1, TRANSFER OF ANCHORAGE FORCES INTO DIAPHRAGM
Replace with the following:
12.14.7.5.1 Transfer of Anchorage Forces into Diaphragm. Diaphragms shall be provided with continuous ties or
struts between diaphragm chords to distribute these anchorage forces into the diaphragm.
EXCEPTION: In buildings with diaphragms of light-framed construction, continuous cross-ties are not required
provided all of the following are satisfied:
1. The unsupported height of the wall does not exceed 12 feet,
2. Anchorages are spaced no more than 4 feet on center,
3. The length of the diaphragm in the direction parallel to the wall being anchored does not exceed 2.5 times the
length of the diaphragm in the orthogonal direction, and
4. The connection extends far enough into the diaphragm to transfer the anchorage force into the diaphragm.
Added chords are permitted to be used to form subdiaphragms to transmit the anchorage forces to the main continuous
cross-ties. The maximum length-to-width ratio of the structural subdiaphragm shall be 2.5 to 1. Connections and
anchorages capable of resisting the prescribed forces shall be provided between the diaphragm and the attached
components. Connections shall extend into the diaphragm a sufficient distance to develop the force transferred into the
diaphragm.
SECTION 12.14.7.5.2, WOOD DIAPHRAGMS
Replace with the following:
12.14.7.5.2 Wood Diaphragms. In wood diaphragms, the continuous ties shall be in addition to the diaphragm
sheathing.
EXCEPTION: Where continuous cross-ties are not required by Section 12.14.7.5.1 and the anchorage connections
extend into the diaphragm a sufficient distance to develop the force transferred into the diaphragm sheathing.
Anchorage shall not be accomplished by use of toenails or nails subject to withdrawal nor shall wood ledgers or framing
be used in cross-gain bending or cross-grain tension. The diaphragm sheathing shall not be considered effective as
providing the ties or struts required by this section.
SECTION 12.14.8.1, SEISMIC BASE SHEAR
Revise, in part, to read as follows:
. . . In calculating SDS, Ss shall be in accordance with Section 11.4.3, but need not be taken larger than 1.5.
[Remainder of section stays the same.]
Commentary to Chapter 12 Modifications
COMMENTARY TO SECTION 12.6
C12.6 ANALYSIS SELECTION PROCEDURE
Table 12.6-1 applies only to buildings without seismic isolation (Chapter 17) or passive energy devices (Chapter 18). The
four basic procedures addressed in Table 12.6-1 are equivalent lateral force (ELF) analysis (Section 12.8), modal response
spectrum (MRS) analysis (Section 12.9), linear response history (LRH) analysis, and nonlinear response history (NRH)
analysis. Requirements for performing response history analysis are provided in Chapter 16. Nonlinear static (pushover)
analysis is not provided as an “approved” analysis procedure in ASCE/SEI 7-05. The value of Ts = SD1/SDS depends on the
site class because SDS and SD1 include such effects. When ELF is not allowed, the analysis must be performed using modal
response spectrum or response history analysis.
ELF is not allowed for buildings with the listed irregularities because it assumes a gradually varying distribution of mass and
stiffness along the height and negligible torsional response. The 3.5Ts limit recognizes that higher modes are more
significant in taller buildings (Lopez and Cruz, 1996; Chopra, 2007), such that the ELF method may underestimate the design
base shear and may not predict correctly the vertical distribution of seismic forces.
Table C12.6-1 demonstrates that 3.5Ts generally increases as ground motion intensity increases and as soils become softer.
Assuming that the fundamental building period is about 0.1 times the number of stories, the maximum building height for
which the ELF applies ranges from about 10 stories for low seismic hazard sites with firm soil to 30 stories for high seismic
hazard sites with soft soil. Since this trend was not intended, the modification to Section 12.6 adds a height limit of 160 feet.
Table C12.6-1 Values of 3.5TS for Various Cities and Various Site Classes
Location
Ss (g)
S1 (g)
3.5Ts (seconds) for Site Class
A&B
C
D
E
Denver
0.219
0.057
0.91
1.29
1.37
1.07
Boston
0.275
0.067
0.85
1.21
1.30
1.03
New York City
0.359
0.070
0.68
0.97
1.08
0.93
Las Vegas
0.582
0.179
1.08
1.50
1.68
1.89
St. Louis
0.590
0.169
1.00
1.40
1.60
1.81
San Diego
1.128
0.479
1.31
1.73
1.99
2.91
Memphis
1.341
0.368
0.96
1.38
1.59
2.25
Charleston
1.414
0.348
0.86
1.25
1.47
2.08
Seattle
1.448
0.489
1.18
1.55
1.78
2.63
San Jose
1.500
0.600
1.40
1.82
2.10
2.12
Salt Lake City
1.672
0.665
1.39
1.81
2.09
3.10
COMMENTARY TO SECTION 12.8.7
C12.8.7 P-delta Limit. ASCE/SEI 7-05 allows amplified forces to be used in a linear elastic analysis where the stability
coefficient, ., exceeds 0.10. By comparison, FEMA 350 requires explicit modeling of P-delta effects for steel momentresisting
frames where . exceeds approximately 0.04. Where the tangent stiffness of the structure may become negative,
dynamic displacement demands can increase significantly (Gupta and Krawinkler, 2000). Structures with . not greater than
0.10 generally are expected to have a positive tangent stiffness, depending on the progression of plastic hinging and strain
hardening. The 2009 Provisions allows structures to exceed this limit only if a nonlinear static analysis including P-delta
effects demonstrates that the tangent stiffness remains positive up to the target displacement computed for the MCER or if
nonlinear dynamic analysis demonstrates adequate resistance to instability.
The occupancy importance factor, I, is inserted into Equation 12.8-16 to correct an error in ASCE/SEI 7. In this way, the
stability coefficient is based on the elastic stiffness of the system.
ADDITIONAL REFERENCES FOR CHAPTER 12 COMMENTARY
Federal Emergency Management Agency. 2000. Recommended Seismic Design Criteria for New Steel Moment-Frame
Buildings, FEMA 350. Prepared for FEMA by the SAC Joint Venture. Federal Emergency Management Agency,
Washington, D.C.
Gupta, A., and H. Krawinkler. 2000. “Dynamic P-delta effects for flexible inelastic steel structures,” ASCE Journal of
Structural Engineering, 126(1):145-154.
Modifications to Chapter 13, Seismic Design Requirements for
Nonstructural Elements
SECTION 13.6.5.5, ADDITIONAL REQUIREMENTS [FOR COMPONENT SUPPORTS]
Replace Item 6f with the following:
Attachments into concrete utilize anchors that have not been prequalified for seismic applications in accordance with
ACI 355.2.
SECTION 13.6.8.2, FIRE PROTECTION SPRINKLER SYSTEMS
IN SEISMIC DESIGN CATEGORY C
Replace with the following:
13.6.8.2 Fire Protection Sprinkler Systems. Fire protection sprinkler systems designed and constructed in accordance
with NFPA 13 shall be deemed to meet the other requirements of this section.
SECTION 13.6.8.3, FIRE PROTECTION SPRINKLER SYSTEMS
IN SEISMIC DESIGN CATEGORIES D THROUGH F
Delete this section and renumber remaining sections.
Commentary to Chapter 13 Modifications
COMMENTARY TO SECTION 13.6.5.5
C13.6.5.5 Additional Requirements [for Component Supports]. As reflected in this section of the standard and in the
footnote to Table 13.6-1, vibration isolated equipment with snubbers is subject to amplified loads as a result of dynamic
impact.
Use of expansion anchors for non-vibration isolated mechanical equipment rated over 10 hp is prohibited based on
experience with older anchor types. The ASCE 7 Seismic Subcommittee is considering a proposal that also would exempt
anchors qualified by simulated seismic testing and long-term vibration testing.
The previous language in Item 6f was intended to identify anchor types that would be considered non-ductile. The previous
requirement has been superseded by requirements for qualification that include checks for ductility and good performance in
earthquake conditions.
COMMENTARY TO SECTION 13.6.8.2
C13.6.8.2 Fire Protection Sprinkler Systems. NFPA 13-2007 applies to Seismic Design Categories C, D, E, and F. The
lateral design procedures of NFPA 13-2007 have been revised for consistency with the ASCE/SEI 7-05 design approach
while retaining traditional sprinkler system design concepts. Using conservative upper-bound values of the various design
parameters, a single lateral force coefficient, Cp, was developed. It is a function of the mapped short period response
parameter Ss. Stresses in the pipe and connections are controlled by limiting the maximum reaction at bracing points as a
function of pipe diameter.
In Seismic Design Category C, the prescriptive requirements of NFPA 13-2007, using a default lateral force of 50 percent of
the weight of the water-filled pipe, provide a conservative design, although application of the NFPA sway bracing calculation
may produce a lower design lateral force.
Page intentionally left blank.
Modifications to Chapter 14, Material Specific Seismic Design and
Detailing Requirements
SECTION 14.1, STEEL
Replace with the following:
14.1 STEEL
Structures, including foundations, constructed of steel to resist seismic loads shall be designed and detailed in accordance
with this standard including the reference documents and additional requirements provided in this section.
14.1.1 Reference Documents. The design, construction, and quality of steel components that resist seismic forces shall
conform to the applicable requirements of the following as amended herein:
1. AISC 360
2. AISC 341
3. AISI NAS
4. AISI S110
5. AISI-GP
6. AISI-PM
7. AISI Lateral
8. AISI WSD
9. ASCE 19
10. ASCE 8
11. SJI Tables
14.1.1.1 Modifications to AISC 341-05. The text of AISC 341 shall be modified as indicated in Sections 14.1.1.1.1
and 14.1.1.1.2. Italics are used for text to indicate requirements that differ from AISC 341.
14.1.1.1.1 Replace Section 15.7 with the following:
15.7 Beam-to-Column Connections
Where a brace or gusset plate connects to both members at a beam-to-column connection, the connection shall
conform to one of the following:
(1) The connection shall accommodate the required rotation at a minimum story drift of 2.5 percent of the story
height or
(2) The connection shall be designed to resist a moment equal to the lesser of the following:
(i) A moment corresponding to 1.1RyFyZ (LRFD) or (1.1/1.5)RyFyZ (ASD), as appropriate, of the beam.
(ii) A moment corresponding to S1.1RyFyZ (LRFD) or S[(1.1/1.5)RyFyZ] (ASD), as appropriate, of the column.
This moment shall be considered in combination with the required strength of the brace connection and beam
connection, including amplified diaphragm collector forces.
1.4.1.1.1.2 Add new Section 16.7 as follows:
16.7 Beam-to-Column Connections
Where a brace or gusset plate connects to both members at a beam-to-column connection, the connection shall
conform to one of the following:
(1) The connection shall accommodate the required rotation at a minimum story drift of 2.5 percent of the story
height or
(2) The connection shall be designed to resist a moment equal to the lesser of the following:
(i) A moment corresponding to 1.1RyFyZ (LRFD) or (1.1/1.5)RyFyZ (ASD), as appropriate, of the beam.
(ii) A moment corresponding to S1.1RyFyZ (LRFD) or S[(1.1/1.5)RyFyZ] (ASD), as appropriate, of the column.
This moment shall be considered in combination with the required strength of the brace connection and beam
connection, including amplified diaphragm collector forces.
14.1.2 Seismic Design Categories B and C. Steel structures assigned to Seismic Design Category B or C shall be of
any construction permitted by the reference documents in Section 14.1.1. An R factor as set forth in Table 12.2-1 is
permitted where the structure is designed and detailed in accordance with the requirements of AISC 341 for structural
steel buildings, AISI S110 for cold-formed steel construction, or AISI Lateral for light-framed cold-formed steel
construction. Systems not detailed in accordance with AISC 341, AISI S110, or AISI Lateral shall use the R factor
designated for “Structural steel systems not specifically detailed for seismic resistance” in Table 12.2-1.
14.1.3 Seismic Design Categories D through F. Steel structures assigned to Seismic Design Category D, E, or F shall
be designed and detailed in accordance with AISC 341 for structural steel, AISI S110 for cold-formed steel construction,
or AISI Lateral for light-framed cold-formed steel construction.
14.1.4 Cold-formed Steel. The design of cold-formed carbon or low-alloy steel to resist seismic loads shall be in
accordance with the requirements of AISI NAS, AISI S110 and the design of cold-formed stainless steel structural
members to resist seismic loads shall be in accordance with the requirements of ASCE 8.
14.1.4.1 Modifications to AISI S110 (2007 edition). The text of AISI S110 shall be modified as indicated in Sections
14.1.4.1.1 through 14.1.2.1.5. Italics are used for text within Sections 14.1.4.1.1 through 14.1.2.1.5 to indicate
requirements that differ from AISI S110.
14.1.4.1.1 AISI S110, Section D1. Revise Section D1 to read as follows:
D1 Cold-Formed Steel Special Bolted Moment Frames (CFS-SBMF)
Cold-formed steel–special bolted moment frames (CFS-SBMF) systems shall withstand inelastic deformations
through friction and bearing at their bolted connections. Beams, columns, and connections shall satisfy the
requirements in this section. CFS-SBMF systems shall be limited to one-story structures, no greater than 35 feet in
height, without column splices and satisfying the requirements in this section. The SBMF shall engage all columns
supporting the roof or floor above. The single size beam and single size column with the same bolted moment
connection detail shall be used for each frame. The frame is to be supported on a level floor or foundation.
14.1.4.1.2 AISI S110, Section D1.1.1. Revise Section D1.1.1 to read as follows:
D1.1.1 Connection Limitations
Beam-to-column connections in CFS-SBMF systems shall be bolted connections with snug-tight highstrength
bolts. The bolt spacing and edge distance shall be in accordance with the limits of AISI S100,
Section E3. The 8-bolt configuration shown in Table D1-1 shall be used. The faying surfaces of the beam
and column in the bolted moment connection region shall be free of any lubricants or debris.
14.1.4.1.3 AISI S110, Section D1.2.1. Revise Section D1.2.1 to read as follows:
D1.2.1 Beam Limitations
In addition to the requirements of Section D1.2.3, beams in CFS-SBMF systems shall be ASTM A653 Gr. 55
galvanized steel cold-formed C-sections members with lips, and designed in accordance with Chapter C of
AISI S100. The beam depth shall be between 12 in (305 mm) and 20 in (508 mm). The flat depth-to-thickness
ratio of the web shall not exceed 6.18 .
14.1.4.1.4 AISI S110, Section D1.2.2. Revise Section D1.2.2 to read as follows:
D1.2.2 Column Limitations
In addition to the requirements of D1.2.3, columns in CFS-SBMF systems shall be ASTM A500 Gr. B coldformed
hollow structural section (HSS) members painted with a standard industrial finished surface, and
designed in accordance with Chapter C of AISI S100. The column depth shall be between 8 in (203 mm) and
12 in (305 mm). The flat depth-to-thickness ratio shall not exceed 1.40 . Equation
Equation Equation
E / Fy
14.1.4.1.5 AISI S110, Section D1.3. Revise Section D1.3 to read as follows:
D1.3 Design Story Drift
Where the applicable building code does not contain design coefficients for CSF-SBMF systems, the
provisions of Appendix 1 shall apply. The design story drift shall not exceed 0.03h, unless approved by
authority having jurisdiction. In no case shall the design story drift exceed 0.05h.
For structures having a period less than TS, as defined in the applicable building code, alternate methods of
computing . shall be permitted, provided such alternate methods are acceptable to the authority having
jurisdiction.
[Remainder of Section 14.1 is unchanged.]
SECTION 14.2.2, MODIFICATIONS TO ACI 318
Replace with the following:
14.2.2 Modifications to ACI 318. The text of ACI 318 shall be modified as indicated in Sections 14.2.2.1 through
14.2.2.9. Italics are used for text within Sections 14.2.2.1 through 14.2.2.9 to indicate provisions that differ from ACI
318.
14.2.2.1 Definitions. Add the following definitions to Section 2.2.
DETAILED PLAIN CONCRETE STRUCTURAL WALL: A wall complying with the requirements of Chapter 22.
ORDINARY PRECAST STRUCTURAL WALL: A precast wall complying with the requirements of Chapters 1
through 18.
WALL PIER: A wall segment with a horizontal length-to-thickness ratio of at least 2.5, but not exceeding 6, whose
clear height is at least two times its horizontal length.
14.2.2.2 ACI 318, Section 7.10. Modify Section 7.10 by revising Section 7.10.5.6 to read as follows:
7.10.5.6 Where anchor bolts are placed in the top of columns or pedestals, the bolts shall be enclosed by lateral
reinforcement that also surrounds at least four vertical bars of the column or pedestal. The lateral reinforcement
shall be distributed within 5 in. of the top of the column or pedestal, and shall consist of at least two No.4 or three
No.3 bars. In structures assigned to Seismic Design Categories C, D, E or F, the ties shall have a hook on each free
end that complies with 7.1.4.
14.2.2.3 Scope. Modify Section 21.1.1.3 to read as follows:
21.1.1.3 All members shall satisfy requirements of Chapters 1 to 19 and 22. Structures assigned to SDC B, C, D, E,
or F also shall satisfy 21.1.1.4 through 21.1.1.8, as applicable, except as modified by the requirements of Chapters
14 and 15 of this document.
14.2.2.4 Intermediate Precast Structural Walls: Modify Section 21.4 by renumbering Section 21.4.3 to Section
21.4.4 and adding new Sections 21.4.3 and 21.4.5, to read as follows:
21.4 Connections that are designed to yield shall be capable of maintaining 80 percent of their design strength at
the deformation induced by design displacement, or shall use type 2 mechanical splices.
21.4.4 Elements of the connection that are not designed to yield shall develop at least 1.5 Sy.
21.4.5 Wall piers not designed as part of a moment frame shall have transverse reinforcement designed to resist the
shear forces determined from Section 21.3.3. Spacing of transverse reinforcement shall not exceed 8 in. Transverse
reinforcement shall be extended beyond the pier clear height for at least 12 in.
EXCEPTIONS: The preceding requirement need not apply in the following situations:
1. Wall piers that satisfy Section 21.13.
2. Wall piers along a wall line within a story where other shear wall segments provide lateral support to the wall
piers and such segments have a total stiffness of at least six times the sum of the in-plane stiffnesses of all the
wall piers.
Wall segments with a horizontal length-to-thickness ratio less than 2.5 shall be designed as columns.
14.2.2.5 Wall Piers and Wall Segments. Modify Section 21.9 by adding a new Section 21.9.10 to read as follows:
21.9.10 Wall Piers and Wall Segments in Special Structural Walls.
21.9.10.1 Wall piers not designed as a part of a special moment-resisting frame shall have transverse reinforcement
designed to satisfy the requirements in Section 21.9.10.2.
EXCEPTIONS:
1. Wall piers that satisfy Section 21.13.
2. Wall piers along a wall line within a story where other shear wall segments provide lateral support to the wall
piers, and such segments have a total stiffness of at least six times the sum of the in-plane stiffnesses of all the
wall piers.
21.9.10.2 Transverse reinforcement with seismic hooks at both ends shall be designed to resist the shear forces
determined from Section 21.6.5.1. Spacing of transverse reinforcement shall not exceed 6 in. (152 mm). Transverse
reinforcement shall be extended beyond the pier clear height for at least 12 in. (304 mm).
21.9.10.3 Wall segments with a horizontal length-to-thickness ratio less than 2.5 shall be designed as columns.
14.2.2.6 Special Precast Structural Walls. Modify Section 21.10.2 to read as follows:
21.10.2 Special structural walls constructed using precast concrete shall satisfy all the requirements of Section 21.9
in addition to 21.4 as modified in Section 14.2.2.7.
14.2.2.7 Foundations. Modify Section 21.12.1.1 to read as follows:
21.12.1.1 Foundations resisting earthquake-induced forces or transferring earthquake-induced forces between
structure and ground shall comply with requirements of Section 21.12 and other applicable code provisions unless
modified by Sections 12.1.5, 12.13 or 14.2 of ASCE/SEI 7-05.
14.2.2.8 Detailed Plain Concrete Shear Walls. Modify Section 22.6 by adding a new Section 22.6.7 to read:
22.6.7 Detailed Plain Concrete Shear Walls.
22.6.7.1 Detailed plain concrete shear walls are walls conforming to the requirements for ordinary plain concrete
shear walls and 22.6.7.2
22.6.7.2 Reinforcement shall be provided as follows:
a. Vertical reinforcement of at least 0.20 in.2 (129 mm2) in cross-sectional area shall be provided continuously
from support to support at each corner, at each side of each opening, and at the ends of walls. The continuous
vertical bar required beside an opening is permitted to substitute for the No. 5 bar required by Section 22.6.6.5.
b. Horizontal reinforcement at least 0.20 in.2 (129 mm2) in cross-sectional area shall be provided:
1. Continuously at structurally connected roof and floor levels and at the top of walls.
2. At the bottom of load-bearing walls or in the top of foundations where doweled to the wall
3. At a maximum spacing of 120 in. (3048 mm).
Reinforcement at the top and bottom of openings, where used in determining the maximum spacing specified in Item
3 in the preceding text, shall be continuous in the wall.
14.2.2.9 Strength Requirements for Anchors: Modify Section D.4 by adding a new exception at the end of Section
D.4.2.2 to read as follows:
EXCEPTION: If Nb is determined using Equation D-7, the concrete breakout strength of Section D.4.2 shall be
considered satisfied by the design procedure of Sections D.5.2 and D.6.2 without the need for testing regardless of
anchor bolt diameter and tensile embedment.
SECTIONS 14.2.3, ADDITIONAL DETAILING REQUIREMENTS FOR CONCRETE PILES,
AND 14.2.3.1, CONCRETE PILE REQUIREMENTS FOR SDC C
Replace with the following:
14.2.3 Additional Detailing Requirements for Concrete Piles. In addition to the foundation requirements set forth in
ACI 318 Sections 12.1.5, 12.13 and 21.12, design, detailing and construction of concrete piles shall conform to the
provisions of this section.
14.2.3.1 Concrete Pile Requirements for Seismic Design Category C. Concrete piles in structures assigned to
Seismic Design Category C shall comply with the requirements of this section.
14.2.3.1.1 Anchorage of Piles. All concrete piles and concrete filled pipe piles shall be connected to the pile cap by
embedding the pile reinforcement in the pile cap for a distance equal to the development length as specified in ACI 318
as modified by Section 14.2.2 of this standard or by the use of field-placed dowels anchored in the concrete pile. For
deformed bars, the development length is the full development length for compression or tension, in the case of uplift,
without reduction in length for excess area.
Hoops, spirals, and ties shall be terminated with seismic hooks as defined in ACI 318 Section 2.2.
Where a minimum length for reinforcement or the extent of closely spaced confinement reinforcement is specified at the
top of the pile, provisions shall be made so that those specified lengths or extents are maintained after pile cut-off.
SECTION 14.2.3.2, CONCRETE PILE REQUIREMENTS FOR SEISMIC DESIGN
CATEGORIES D THROUGH F
Replace Sections 14.2.3.2.1 through 14.2.3.2.5 with the following:
14.2.3.2.1 Site Class E or F Soil. Where concrete piles are used in Site Class E or F, they shall have transverse
reinforcement in accordance with ACI 318 Sections 21.6.4.2 through 21.6.4.4 within seven pile diameters of the pile cap
and the interfaces between strata that are hard or stiff and strata that are liquefiable or are composed of soft to medium
stiff clay.
14.2.3.2.2 Nonapplicable ACI 318 Sections for Grade Beam and Piles. ACI 318 Section 21.12.3.3 need not apply
where grade beams have the required strength to resist the forces from the load combinations with overstrength factor of
Section 12.4.3.2 or 12.14.3.2.2. ACI 318 Section 21.12.4.4(a) need not apply to concrete piles, and Section 21.12.4.4(b)
need not apply to precast, prestressed concrete piles.
14.2.3.2.3 Reinforcement for Uncased Concrete Piles (SDC D through F). Reinforcement shall be provided where
required by analysis. For uncased cast-in-place drilled or augered concrete piles, a minimum of four longitudinal bars
with a minimum longitudinal reinforcement ratio of 0.005 and transverse reinforcement in accordance with ACI 318
Sections 21.6.4.2 through 21.6.4.4 shall be provided throughout the minimum reinforced length of the pile as defined
below starting at the top of the pile. The longitudinal reinforcement shall extend beyond the minimum reinforced length
of the pile by the tension development length.
The minimum reinforced length of the pile shall be taken as the greater of:
1. One-half of the pile length;
2. A distance of 10 ft (3 m);
3. Three times the pile diameter;
4. The flexural length of the pile which shall be taken as the length of from the bottom of the pile cap to a point where
the concrete section cracking moment multiplied by a resistance factor 0.4 exceeds the required factored moment at
that point.
In addition, for piles located in Site Class E or F, longitudinal reinforcement and transverse confinement reinforcement,
as described above, shall extend the full length of the pile.
Where transverse reinforcement is required, transverse reinforcing ties shall be a minimum of No. 3 bars for up to 20-in.-
diameter (300 mm) piles and No.4 bars for piles of larger diameter.
In Site Classes A through D, longitudinal reinforcement and transverse confinement reinforcement, as defined above,
shall extend a minimum of seven times the pile diameter above and below the interfaces of soft to medium stiff clay or
liquefiable strata except that transverse reinforcing ties not located within the minimum reinforced length shall be
permitted to use a transverse spiral reinforcement ratio of not less than one-half of that required in ACI 318 Section
21.6.4.4(a). Spacing of transverse reinforcement not located within the minimum reinforced length is permitted to be
increased, but shall not exceed the least of the following:
1. 12 longitudinal bar diameters;
2. One-half the pile diameter;
3. 12 in. (305 mm).
14.2.3.2.4 Reinforcement for Metal-Cased Concrete Piles (SDC D through F). Reinforcement requirements are the
same as for uncased concrete piles.
EXCEPTION: Spiral-welded metal-casing of a thickness not less than No. 14 gauge can be considered as
providing concrete confinement equivalent to the closed ties or equivalent spirals required in an uncased
concrete pile, provided that the metal casing is adequately protected against possible deleterious action due
to soil constituents, changing water levels, or other factors indicated by boring records of site conditions.
14.2.3.2.5 Reinforcement for Precast Concrete Piles (SDC D through F). Transverse confinement reinforcement
consisting of closed ties or equivalent spirals shall be provided in accordance with ACI 318 Sections 21.6.4.2 through
21.6.4.4 for the full length of the pile.
EXCEPTION: In other than Site Classes E or F, the specified transverse confinement reinforcement shall be provided
within three pile diameters below the bottom of the pile cap, but it shall be permitted to use a transverse reinforcing ratio
of not less than one-half of that required in ACI 318 Section 21.6.4.4(a) throughout the remainder of the pile length.
[Remainder of Section 14.2.3.2 is unchanged.]
NEW SECTION 14.2.4, ACCEPTANCE CRITERIA FOR SPECIAL PRECAST STRUCTURAL
WALLS BASED ON VALIDATION TESTING
Add the following new section:
14.2.4 Acceptance Criteria for Special Precast Structural Walls Based on Validation Testing
14.2.4.1 Notation
Symbols additional to those in ACI 318 Chapter 2 are defined.
Emax = maximum lateral resistance of test module determined from test results (forces or moments)
En = nominal lateral resistance of test module calculated using specified geometric properties of test members,
specified yield strength of reinforcement, specified compressive strength of concrete, a strain compatibility
analysis or deformation compatibility analysis for flexural strength and a strength reduction factor f of 1.0
Ent = calculated lateral resistance of test module using the actual geometric properties of test members, the actual
strengths of reinforcement, concrete, and coupling devices, obtained by testing per Sections 14.2.4.7.7,
14.2.4.7.8, and 14.2.4.7.9, and a strength reduction factor f of 1.0
. = drift ratio
ß = relative energy dissipation ratio
14.2.4.2 Definitions
Definitions additional to those in ACI 318 Chapter 2 are defined.
14.2.4.2.1 Coupling Elements. Devices or beams connecting adjacent vertical boundaries of structural walls and used
to provide stiffness and energy dissipation for the connected assembly greater than the sum of those provided by the
connected walls acting as separate units.
14.2.4.2.2 Drift Ratio. Total lateral deformation of the test module divided by the height of the test module.
14.2.4.2.3 Global Toughness. The ability of the entire lateral force resisting system of the prototype structure to
maintain structural integrity and continue to carry the required gravity load at the maximum lateral displacements
anticipated for the ground motions of the maximum considered earthquake.
14.2.4.2.4 Prototype Structure. The concrete wall structure for which acceptance is sought.
14.2.4.2.5 Relative Energy Dissipation Ratio. Ratio of actual to ideal energy dissipated by test module during
reversed cyclic response between given drift ratio limits, expressed as the ratio of the area of the hysteresis loop for that
cycle to the area of the circumscribing parallelograms defined by the initial stiffnesses during the first cycle and the peak
resistances during the cycle for which the relative energy dissipation ratio is calculated. See Section 14.2.4.9.1.3.
14.2.4.2.5 Test Module. Laboratory specimen representing the critical walls of the prototype structure. See Section
14.2.4.5.
14.2.4.3 Scope and General Requirements
14.2.4.3.1 These provisions define minimum acceptance criteria for new precast structural walls, including coupled
precast structural walls, designed for regions of high seismic risk or for structures assigned to high seismic performance
or design categories, where acceptance is based on experimental evidence and mathematical analysis.
14.2.4.3.2 These provisions are applicable to precast structural walls, coupled or uncoupled, with height to length, hw/lw,
ratios equal to or greater than 0.5. These provisions are applicable for either prequalifying precast structural walls for a
specific structure or prequalifying a new precast wall type for construction in general.
14.2.4.3.3 Precast structural walls shall be deemed to have a response that is at least equivalent to the response of
monolithic structural walls designed in accordance with ACI 318 Sections 21.1 and 21.9, and the corresponding
structural walls of the prototype structure shall be deemed acceptable, when all of the conditions in Sections 14.2.4.3.3.1
through 14.2.4.3.3.5 are satisfied.
14.2.4.3.3.1 The prototype structure satisfies all applicable requirements of these provisions and of ACI 318 except
Section 21.9.
14.2.4.3.3.2 Tests on wall modules satisfy the conditions in Sections 14.2.4.4 and 14.2.4.9.
14.2.4.3.3.3 The prototype structure is designed using the design procedure substantiated by the testing program.
14.2.4.3.3.4 The prototype structure is designed and analyzed using effective initial properties consistent with those
determined in accordance with Section 14.2.4.7.11, and the prototype structure meets the drift limits of these provisions.
14.2.4.3.3.5 The structure as a whole, based on the results of the tests of Section 14.2.4.3.3.2 and analysis, is
demonstrated to have adequate global toughness (the ability to retain its structural integrity and support its specified
gravity loads) through peak displacements equal to or exceeding the story-drift ratios specified in Section 14.2.4.7.4,
14.2.4.7.5 or 14.2.4.7.6, as appropriate.
14.2.4.4 Design Procedure
14.2.4.4.1 Prior to testing, a design procedure shall be developed for the prototype structure and its walls. That
procedure shall account for effects of material non-linearity, including cracking, deformations of members and
connections, and reversed cyclic loading. The design procedure shall include the procedures specified in Sections
14.2.4.4.1.1 through 14.2.4.4.1.4 and shall be applicable to all precast structural walls, coupled and uncoupled, of the
prototype structure.
14.2.4.4.1.1 Procedures shall be specified for calculating the effective initial stiffness of the precast structural walls, and
of coupled structural walls, that are applicable to all the walls of the prototype structure.
14.2.4.4.1.2 Procedures shall be specified for calculating the lateral strength of the precast structural walls, and of
coupled structural walls, applicable to all precast walls of the prototype structure.
14.2.4.4.1.3 Procedures shall be specified for designing and detailing the precast structural walls so that they have
adequate ductility capacity. These procedures shall cover wall shear strength, sliding shear strength, boundary tie
spacing to prevent bar buckling, concrete confinement, reinforcement strain, and any other actions or elements of the
wall system that can affect ductility capacity.
14.2.4.4.1.4 Procedures shall be specified for determining that an undesirable mechanism of nonlinear response, such as
a story mechanism due to local buckling of the reinforcement or splice failure, or overall instability of the wall, does not
occur.
14.2.4.4.2 The design procedure shall be used to design the test modules and shall be documented in the test report.
14.2.4.4.3 The design procedure used to proportion the test specimens shall define the mechanism by which the system
resists gravity and earthquake effects and shall establish acceptance values for sustaining that mechanism. Portions of
the mechanism that deviate from code requirements shall be contained in the test specimens and shall be tested to
determine acceptance values.
14.2.4.5 Test Modules
14.2.4.5.1 At least two modules shall be tested. At least one module shall be tested for each limiting engineering design
criteria (shear, axial load and flexure) for each characteristic configuration of precast structural walls, including
intersecting structural walls or coupled structural walls. If all the precast walls of the structure have the same
configuration and the same limiting engineering design criterion, then two modules shall be tested. Where intersecting
precast wall systems are to be used, the response for the two orthogonal directions shall be tested.
14.2.4.5.2 Where the design requires the use of coupling elements, those elements shall be included as part of the test
module.
14.2.4.5.3 Modules shall have a scale large enough to represent the complexities and behavior of the real materials and
of the load transfer mechanisms in the prototype walls and their coupling elements, if any. Modules shall have a scale
not less than one half and shall be full-scale if the validation testing has not been preceded by an extensive analytical and
experimental development program in which critical details of connections are tested at full scale.
14.2.4.5.4 The geometry, reinforcing details, and materials properties of the walls, connections, and coupling elements
shall be representative of those to be used in the prototype structure.
14.2.4.5.5 Walls shall be at least two panels high unless the prototype structure is one for which a single panel is to be
used for the full height of the wall.
14.2.4.5.6 Where precast walls are to be used for bearing wall structures, as defined in ASCE/SEI 7-05, the test modules
shall be subject during lateral loading to an axial load stress representative of that anticipated at the base of the wall in
the prototype structure.
14.2.4.5.7 The geometry, reinforcing, and details used to connect the precast walls to the foundation shall replicate those
to be used in the prototype structure.
14.2.4.5.8 Foundations used to support the test modules shall have geometric characteristics, and shall be reinforced and
supported, so that their deformations and cracking do not affect the performance of the modules in a way that would be
different than in the prototype structure.
14.2.4.6 Testing Agency. Testing shall be carried out by an independent testing agency approved by the Authority
Having Jurisdiction. The testing agency shall perform its work under the supervision of a registered design professional
experienced in seismic structural design.
14.2.4.7 Test Method
14.2.4.7.1 Test modules shall be subjected to a sequence of displacement-controlled cycles representative of the drifts
expected under earthquake motions for the prototype structure. If the module consists of coupled walls, approximately
equal drifts (within 5 percent of each other) shall be applied to the top of each wall and at each floor level. Cycles shall
be to predetermined drift ratios as defined in Sections 14.2.4.7.2 through 14.2.4.7.6.
14.2.4.7.2 Three fully reversed cycles shall be applied at each drift ratio.
14.2.4.7.3 The initial drift ratio shall be within the essentially linear elastic response range for the module. See
14.2.4.7.11. Subsequent drift ratios shall be to values not less than 5/4 times, and not more than 3/2 times, the previous
drift ratio.
14.2.4.7.4 For uncoupled walls, testing shall continue with gradually increasing drift ratios until the drift ratio in percent
equals or exceeds the larger of : (a) 1.5 times the drift ratio corresponding to the design displacement or (b) the
following value:
(14.2.4-1)
where hw = height of entire wall for prototype structure (in inches) and lw = length of entire wall in direction of shear
force (in inches).
14.2.4.7.5 For coupled walls, hw/lw in Equation 14.2.4-1 shall be taken as the smallest value of hw/lw for any individual
wall of the prototype structure.
14.2.4.7.6 Validation by testing to limiting drift ratios less than those given by Equation 14.2.4-1 shall be acceptable
provided testing is conducted in accordance with this document to drift ratios equal or exceeding of those determined for
the response to a suite of nonlinear time history analyses conducted in accordance with the 2009 NEHRP Recommended
Seismic Provisions for risk-targeted maximum considered earthquake ground motions.
14.2.4.7.7 Actual yield strength of steel reinforcement shall be obtained by testing coupons taken from the same
reinforcement batch as used in the test module. Two tests, conforming to the ASTM specifications cited in ACI 318
Section 3.8, shall be made for each reinforcement type and size. Equation
Equation
14.2.4.7.8 Actual compressive strength of concrete shall be determined by testing of concrete cylinders cured under the
same conditions as the test module and tested at the time of testing the module. Testing shall conform to the applicable
requirements of ACI 318 Sections 5.6.1 through 5.6.4.
14.2.4.7.9 Where strength and deformation capacity of coupling devices does not depend on reinforcement tested as
required in Section 14.2.4.7.7, the effective yield strength and deformation capacity of coupling devices shall be obtained
by testing independent of the module testing.
14.2.4.7.10 Data shall be recorded from all tests such that a quantitative interpretation can be made of the performance
of the modules. A continuous record shall be made of test module drift ratio versus applied lateral force, and
photographs shall be taken that show the condition of the test module at the peak displacement and after each key testing
cycle.
14.2.4.7.11 The effective initial stiffness of the test module shall be calculated based on test cycles to a force between
0.6Ent and 0.9Ent, and using the deformation at the strength of 0.75Ent to establish the stiffness.
14.2.4.8 Test Report
14.2.4.8.1 The test report shall contain sufficient evidence for an independent evaluation of all test procedures, design
assumptions, and the performance of the test modules. As a minimum, all of the information required by Sections
14.2.4.8.1.1 through 14.2.4.8.1.11 shall be provided.
14.2.4.8.1.1 A description shall be provided of the design procedure and theory used to predict test module strength,
specifically the test module nominal lateral resistance, En, and the test module actual lateral resistance Ent.
14.2.4.8.1.2 Details shall be provided of test module design and construction, including fully dimensioned engineering
drawings that show all components of the test specimen.
14.2.4.8.1.3 Details shall be provided of specified material properties used for design, and actual material properties
obtained by testing in accordance with Section 14.2.4.7.7.
14.2.4.8.1.4 A description shall be provided of test setup, including fully dimensioned diagrams and photographs.
14.2.4.8.1.5 A description shall be provided of instrumentation, its locations, and its purpose.
14.2.4.8.1.6 A description and graphical presentation shall be provided of applied drift ratio sequence.
14.2.4.8.1.7 A description shall be provided of observed performance, including photographic documentation, of the
condition of each test module at key drift ratios including, (as applicable), the ratios corresponding to first flexural
cracking or joint opening, first shear cracking, and first crushing of the concrete for both positive and negative loading
directions, and any other significant damage events that occur. Photos shall be taken at peak drifts and after the release
of load.
14.2.4.8.1.8 A graphical presentation shall be provided of lateral force versus drift ratio response.
14.2.4.8.1.9 A graphical presentation shall be provided of relative energy dissipation ratio versus drift ratio.
14.2.4.8.1.10 A calculation shall be provided of effective initial stiffness for each test module as observed in the test and
as determined in accordance with Section 14.2.4.7.11 and a comparison made as to how accurately the design procedure
has been able to predict the measured stiffness. The design procedure shall be used to predict the overall structural
response and a comparison made as to how accurately that procedure has been able to predict the measured response.
14.2.4.8.1.11 The test date, report date, name of testing agency, report author(s), supervising registered design
professional, and test sponsor shall be provided.
14.2.4.9 Test Module Acceptance Criteria
14.2.4.9.1 The test module shall be deemed to have performed satisfactorily when all of the criteria Sections 14.2.4.9.1.1
through 14.2.4.9.1.3 are met for both directions of in-plane response. If any test module fails to pass the validation
testing required by these provisions for any test direction, then the wall system has failed the validation testing.
14.2.4.9.1.1 Peak lateral strength obtained shall be at least 0.9Ent and not greater than 1.2 Ent.
14.2.4.9.1.2 In cycling up to the drift level given by Sections 14.2.4.7.4 through 14.2.4.7.6, fracture of reinforcement or
coupling elements, or other significant strength degradation, shall not occur. For a given direction, peak lateral strength
during any cycle of testing to increasing displacement shall not be less than 0.8 times Emax for that direction.
14.2.4.9.1.3 For cycling at the given drift level for which acceptance is sought in accordance with Section 14.2.4.7.4,
14.2.4.7.5 or 14.2.4.7.6, as applicable, the parameters describing the third complete cycle shall have satisfied the
following:
1. The relative energy dissipation ratio shall not be less than 1/8 and
2. The secant stiffness between drift ratios of -1/10 and +1/10 of the maximum applied drift shall not be less than 0.10
times the stiffness for the initial drift ratio specified in Section 14.2.4.7.3.
SECTION 14.4.5, MODIFICATIONS TO CHAPTER 1 OF ACI530/ASCE 5/TMS 402
Add the following new sections:
14.4.5.3 Plain (unreinforced) AAC masonry shear walls shall satisfy the requirements of Section 1.14.2.2.6 of ACI
530/ASCE 5/TMS 402.
14.4.5.4 Ordinary reinforced AAC masonry shear walls shall satisfy the requirements of Section 1.14.2.2.8 of ACI
530/ASCE 5/TMS 402.
Commentary to Chapter 14 Modifications
COMMENTARY TO SECTION 14.1.1
ASCE/SEI 7-05 included two different systems for both eccentrically braced frames (EBF) and buckling restrained braced
frames (BRBF). The primary distinction between these two systems was whether or not there were moment resisting beamcolumn
connections within the braced bays. However, testing at the University of California at Berkeley (Uriz and Mahin,
2004) has indicated designs that do not properly account for the stiffness and distribution of forces in braced frame
connections may be subject to undesirable performance. This modification to ASCE/SEI 7-05 consolidates the EBF and
BRBF building frame systems into a single designation with proper consideration of the beam-column connection demands.
This modification to ASCE/SEI 7-05 and the related changes to AISC 341-05 Sections 15.7 and 16.7 also allow the engineer
either to:
1. Provide a fully restrained moment connection meeting the requirements for ordinary moment connections in AISC 341-
05 and thereby directly providing a load path to resist the connection force and deformation demands or
2. Provide a connection with the ability to accommodate the potential rotation demands. An example of this would be a
configuration tested at Lehigh University (Figure 1 of Fahnestock, et. al. 2006) that effectively formed a pinned
condition in the beam just beyond the beam-column-brace connection.
COMMENTARY TO SECTION 14.1.4
C14.1.4 Cold-Formed Steel. This section adopts three standards by direct reference: AISI NAS, North American
Specification for the Design of Cold-Formed Steel Structural Members, AISI S110, Standard for Seismic Design of Cold-
Formed Steel Structural Systems – Special Bolted Moment Frames, and ASCE/SEI 8, Specification for the Design of Cold
Formed Stainless Steel Structural Members.
Each document has specific limits of applicability. AISI NAS applies to the design of structural members that are coldformed
to shape from carbon or low-alloy steel sheet, strip, plate or bar not more than one-inch in thickness (AISI NAS,
Section A1.1). Building on the requirements of AISI NAS, AISI S110 has additional special seismic design provisions for a
newly designated seismic force resisting system entitled “cold-formed steel – special bolted moment frame (CFS-SBMF).”
Finally, ASCE 8 governs the design of structural members that are cold-formed to shape from annealed and cold-rolled sheet,
strip, plate, or flat bar stainless steels (ASCE 8, Section 1.1.1). All three documents focus on load-carrying members in
buildings; however, allowances are made for applications in nonbuilding structures, if dynamic effects are appropriately
considered.
Within AISI NAS and ASCE 8, there are requirements on the general provisions for the applicable types of steel; design of
elements, members, structural assemblies, connections and joints; and mandatory testing. In addition, AISI NAS contains a
chapter on the design of cold-formed steel structural members and connections undergoing cyclic loading. Both standards
contain extensive commentaries for the benefit of the user.
C14.1.4.1.1 CFS-SBMF need to use the same-size beams and same-size columns throughout. In addition, the system needs
to engage all primary columns, which support the roof or floor above, and those columns need to be supported on a level
floor or foundation.
C14.1.4.1.2 These modifications were made for consistency with the test database.
C14.1.4.1.3 To be consistent with the test database (Uang and Sato, 2007), the limitations on both beam depth, steel grade,
and surface treatment are added in Section D1.2.1 of AISI S110.
C14.1.4.1.4 To be consistent with the test database (Uang and Sato, 2007), the limitations on column depth, steel grade, and
surface treatment are added in Section D1.2.2 of AISI S110. The width-thickness ratio was reduced based upon further
review of the test specimens.
C14.1.4.1.5 AISI S110 is intended primarily for industrial platforms; however, the standard is not limited to these nonbuilding
structures and does not prohibit architectural attachments (such as partition walls). Therefore, the 0.05h drift limit in
Section D1.3 of AISI S110 has been reduced to 0.03h to more closely align with the 0.025h drift limit of ASCE/SEI 7. The
sentence, “In no case shall the design story drift exceed 0.05h.” was added to ensure an absolute upper bound on the drift
limit.
C14.1.4.2 Light-Framed Cold-Formed Construction. This subsection of cold-formed steel relates to light-framed
construction, which is defined as a method of construction where the structural assemblies are formed primarily by a system
of repetitive wood or cold-formed steel framing members or subassemblies of these members (ASCE/SEI 7, Section 11.2).
Not only does this subsection repeat the direct adoptions of AISI NAS and ASCE 8, but it also allows the user to choose from
an additional suite of standards that address different aspects of construction, including the following:
1. AISI GP, Standard for Cold-Formed Steel Framing – General Provisions, applies to the design, construction, and
installation of structural and non-structural cold-formed steel framing members where the specified minimum base metal
thickness is between 18 mils and 118 mils (AISI GP, Section A1).
2. AISI WSD, Standard for Cold-Formed Steel Framing – Wall Stud Design, applies to the design and installation of coldformed
steel studs for both structural and nonstructural walls in buildings (AISI WSD, Section A1).
COMMENTARY TO SECTION 14.2
C14.2 CONCRETE
The section adopts by reference ACI 318 for structural concrete design and construction. In addition, modifications to ACI
318 are made that are needed to coordinate the provisions of that material design standard with the provisions of ASCE/SEI
7. Work is ongoing to better coordinate the provisions of the two documents (ACI 318 and ASCE/SEI 7) such that the
provisions in Section 14.2 will be significantly reduced in future editions of ASCE/SEI 7.
C14.2.2.2 ACI 318 Section 7.10. ACI 318 Section 7.10.5.6 prescribes reinforcement details for ties in compression
members. Those details are appropriate for SDC A and B structures. This modification prescribes additional details for ties
around anchor bolts in structures assigned to SDC C through F.
A wall pier is recognized as a separate category of structural element in this document but not ACI 318.
C14.2.2.3 Scope. This provision describes how the ACI 318 provisions should be interpreted for consistency with the
ASCE/SEI 7 provisions.
C14.2.2.4 Intermediate Precast Structural Walls. ACI 318 Section 21.4 imposes requirements on precast walls for
moderate seismic risk applications. Ductile behavior is to be ensured by yielding of the steel elements or reinforcement
between panels or this provision requires the designer to determine the deformation in the connection corresponding to the
earthquake design displacement, and then to check from experimental data that the connection type used can accommodate
that deformation without significant strength degradation.
The wall pier requirements of Section 21.4.5 are patterned after the same requirements of Section 14.2.2.4 for wall piers that
are part of structures in high seismic design categories. The 2006 Edition of the International Building Code restricts yielding
to steel reinforcement only because of concern that steel elements in the body of a connection could fracture due to inelastic
strain demands.
Several steel element connections have been tested under simulated seismic loading and the adequacy of their loaddeformation
characteristics and strain capacity have been demonstrated (Schultz and Magana, 1996). One such connection
was used in the five-story building test that was part of the PRESSS Phase 3 research. The connection was used to provide
damping and energy dissipation, and demonstrated a very large strain capacity (Nakaki et al., 2001). Since then, several other
steel element connections have been developed that can achieve similar results (Banks and Stanton), (Nakaki et al.). In view
of these results, it is appropriate to allow yielding in steel elements that have been shown experimentally to have adequate
strain capacity to maintain at least 80 percent of their yield force through the full design displacement of the structure.
C14.2.2.5 Wall Piers and Wall Segments. Wall piers are typically segments between openings in walls that are thin in the
direction normal to the horizontal length of the wall. In current practice these elements are often not regarded as columns or
as part of the structural walls. If not properly reinforced these elements are vulnerable to shear failure and that failure
prevents the wall from developing the assumed flexural hinging. Section 21.9.10 is written to reduce the likelihood of a
shear failure. Wall segments with a horizontal length-to-thickness ratio less than 2.5 are required to be designed as columns
in compliance with Section 21.9 if they are utilized as part of the lateral-force-resisting system, even though the shortest
cross-sectional dimension may be less than 12 in. in violation of Section 21.6.1.1. Such wall segments may be designed to
comply with Section 21.13 if they are not utilized as part of the lateral-force-resisting system. Wall segments with a
horizontal length-to-thickness ratio larger than or equal to 2.5, which do not meet the definition of wall piers (Section
14.2.2.2), must be designed as special structural walls or as portions of special structural walls in full compliance with
Section 21.9 or 21.10.
C14.2.2.7 Foundations. The intention is that there should be no conflicts between the provisions of ACI 318 Section 21.12
and ASCE/SEI 7-05 Section 12.1.5, 12.13, or 14.2. However, the additional detailing requirements for concrete piles of
Section 14.2.3 can result in conflicts with ACI 318 provisions if the pile in not fully embedded in the soil.
C14.2.2.8 Detailed Plain Concrete Walls. Design requirements for plain masonry walls have existed for many years and
the corresponding type of concrete construction is the plain concrete wall. To allow the use of such walls as the lateral-forceresisting
system in SDC A and B, this provision requires such walls to contain at least the minimal reinforcement specified in
Section 22.6.7.2.
C14.2.2.9 Strength Requirements for Anchors. ACI 318 requires laboratory testing to establish the strength of anchor
bolts greater than 2 in. in diameter or exceeding 25 in. in tensile embedment depth. This modification makes the ACI 318
equation giving the basic concrete breakout strength of a single anchor in tension in cracked concrete applicable irrespective
of the anchor bolt diameter and tensile embedment depth.
Korean Power Engineering (KPE) has made tension tests on anchors with diameters up to 4.25 in. and embedment depths up
to 45 in. and found that the diameter and embedment depth limits of ACI 318 Section D.4.2.2 for the design procedure for
anchors in tension (Section D.5.2) can be eliminated. KPE also has conducted shear tests on anchors with diameters up to 3.0
in. and embedment depths as large as 30 in. and found no effect of the embedment depth on shear strength. However, the
diameter tests showed that the basic shear breakout strength Equation D-24 needed some modification for the complete
elimination of the 2 in. limit to be fully appropriate. Analytical work performed at the University of Stuttgart supports the
need for some modification to Equation D-24. Changes consistent with the Korean and Stuttgart findings have already been
made to the FIB Design Guide for anchors.
COMMENTARY TO SECTION 14.2.3
C14.2.3 Additional Detailing Requirements for Concrete Piles. Chapter 20 of the PCI Bridge Design Manual provides
detailed information on the structural design of piles and on pile to cap connections for precast prestressed concrete piles.
ACI 318 does not contain provisions governing the design and installation of portions of concrete piles, drilled piers, and
caissons embedded in ground except for SDC D, E and F structures.
C14.2.3.1.2 Reinforcement for Uncased Concrete Piles (SDC C). The transverse reinforcing requirements in the
potential plastic hinge zone of uncased concrete piles in Seismic Design Category C is a selective composite of two ACI 318
requirements. In the potential plastic hinge region of an intermediate moment-resisting concrete frame column, the
transverse reinforcement spacing is restricted to the least of: (a) 8 times the diameter of the smallest longitudinal bar, (b) 24
times the diameter of the tie bar, (c) one-half the smallest cross-sectional dimension of the column, and (d) 12 in. Outside of
the potential plastic hinge region of a special moment-resisting frame column, the transverse reinforcement spacing is
restricted to the smaller of 6 times the diameter of the longitudinal column bars and 6 in.
C14.2.3.1.5 Reinforcement for Precast Nonprestressed Concrete Piles (SDC C). Transverse reinforcement requirements
in and outside of the plastic hinge zone of precast nonprestressed piles are clarified. The transverse reinforcement
requirement in the potential plastic hinge zone is a composite of two ACI 318 requirements (see Section C14.2.3.1.2).
Outside of the potential plastic hinge region, the transverse reinforcement spacing is restricted to sixteen (16) times the
longitudinal bar diameter. This should permit the longitudinal bars to reach compression yield before buckling. The
maximum 8-in. tie spacing comes from current building code provisions for precast concrete piles.
C14.2.3.1.6 Reinforcement for Precast Prestressed Piles (SDC C). The transverse and longitudinal reinforcing
requirements given in ACI 318, Chapter 21, were never intended for slender precast prestressed concrete elements and will
result in unbuildable piles. These requirements are based on the Recommended Practice for Design, Manufacture and
Installation of Prestressed Concrete Piling (PCI Committee on Prestressed Concrete Piling, 1993).
Equation 14.2.4-1, originally from ACI 318, has always been intended to be a lower-bound spiral reinforcement ratio for
larger diameter columns. It is independent of the member section properties and can therefore be applied to large or small
diameter piles. For cast-in-place concrete piles and precast prestressed concrete piles, the resulting spiral reinforcing ratios
from this formula are considered to be sufficient to provide moderate ductility capacities (Fanous et al., 2007).
Full confinement per Equation 14.2.4-1 is required for the upper 20 feet of the pile length where curvatures are large. The
amount is relaxed by 50 percent outside of that length in view of lower curvatures and in consideration of confinement
provided by the soil.
C14.2.3.2.3 Reinforcement for Uncased Concrete Piles (SDC D through F). The reinforcement requirements for uncased
concrete piles are taken from the 2006 IBC requirements, and should be adequate to provide ductility in the potential plastic
hinge zones (Fanous et al., 2007).
C14.2.3.2.5 Reinforcement for Precast Concrete Piles (SDC D through F). The transverse reinforcement requirements
for precast nonprestressed concrete piles are taken from the 2006 IBC requirements and are should be adequate to provide
ductility in the potential plastic hinge zones (Fanous et al., 2007).
C14.2.3.2.6 Reinforcement for Precast-Prestressed Piles (SDC D through F). The reduced amounts of transverse
reinforcement specified in this provision compared to those required for column members in ACI 318 are justified by the
results of the study by Fanous et al., 2007. The last paragraph of the provision provides minimum transverse reinforcement
requirements outside of the zone of prescribed ductile detailing.
COMMENTARY TO SECTION 14.2.4
C14.2.4 Acceptance Criteria for Special Precast Structural Walls Based on Validation Testing
C14.2.4.1 Notation. Symbols additional to those in ACI 318 Chapter 2 are defined:
Ah = area of hysteresis loop.
E1,E2 = peak lateral resistance for positive and negative loading, respectively, for third cycle of loading sequence.
f1 = live load factor defined in Section 14.2.4.2.3.
hw = height of column of test module, in. or mm.
K, K’ = initial stiffness for positive and negative loading, respectively, for first cycle.
.1,.2 = drift ratios at peak lateral resistance for positive and negative loading, respectively, for third cycle of loading
sequence.
.1',.2' = drift ratios for zero lateral load for unloading at stiffness K, K’ from peak positive and negative lateral resistance,
respectively, for third cycle of loading sequence.
. = lateral displacement, in. or mm. See Figures. C14.2.4.2.2-1, C14.2.4.2.2-2, and C14.2.4.2.2-3.
.a = allowable story drift, in. or mm. See Table 12.12-1 of ASCE/SEI 7-05.
C14.2.4.2 Definitions
C14.2.4.2.1 Coupling elements. Coupling elements are connections provided at specific intervals along the vertical
boundaries of adjacent structural walls. Coupled structural walls are stiffer and stronger than the same walls acting
independently. For cast-in-place construction effective coupling elements are typically coupling beams having small span-todepth
ratios. The inelastic behavior of such beams is normally controlled by their shear strength. For precast construction,
effective coupling elements can be precast beams connected to the adjacent structural walls either by post-tensioning, ductile
mechanical devices, or grouted-in-place reinforcing bars. The resultant coupled construction can be either emulative of castin-
place construction or non-emulative (jointed). However, for precast construction coupling beams can also be omitted and
mechanical devices used to connect directly the vertical boundaries of adjacent structural walls.
C14.2.4.2.2 Drift ratio. The definition of the drift ratio, ., is illustrated in Figure C14.2.4.2.2-1 for a three panel wall
module. The position of the module at the start of testing, with only its self-weight acting, is indicated by broken lines. The
module is set on a horizontal foundation support that is centered at A and is acted on by a lateral force H applied at the top of
the wall. The self-weight of the wall is distributed uniformly to the foundation support. However, under lateral loading, that
self-weight and any axial gravity load acting at the top of the wall cause overturning moments on the wall that are additional
to the overturning moment Hhw and can affect deformations. The chord AB of the centroidal axis of the wall is the vertical
reference line for drift measurements.
For acceptance testing a lateral force H is applied to the wall through the pin at B. Depending on the geometric and
reinforcement characteristics of the module that force can result in the module taking up any one, or a combination, of the
deformed shapes indicated by solid lines in Figures C14.2.4.2.2-1, C14.2.4.2.2-2 and C14.2.4.2.2-3.
Figure C14.2.4.2.2-2 illustrates several possible components of the displacement . for a wall that is effectively solid while
Figure C14.2.4.2.2-3 illustrates two possibly undesirable components of the displacement .. Regardless of the mode of
deformation of the wall, the lateral force causes the wall at B to displace horizontally by an amount .. The drift ratio is the
angular rotation of the wall chord with respect to the vertical and for the setup shown equals . / hw where hw is the wall
height and is equal to the distance between the foundation support at A and the load point at B. Where prestressing steel is
used in wall members, the stress fps in the reinforcement at the nominal and the probable lateral resistance shall be calculated
in accordance with ACI 318 Section 18.7.
C14.2.4.2.3 Global toughness. These provisions describe acceptance criteria for special precast structural walls based on
validation testing. The requirements of Section 21.1.1.8 of ACI 318 concerning toughness cover both to the energy
dissipation of the wall system which, for monolithic construction, is affected primarily by local plastic hinging behavior and
the toughness of the prototype structure as a whole. The latter is termed “global toughness” in these provisions and is a
condition that does not apply to the walls alone. That global toughness requirement can be satisfied only though analysis of
the performance of the prototype structure as a whole when the walls perform to the criteria specified in these provisions.
The required gravity load for global toughness evaluations is the value given by these provisions. For conformity with
Section 9.2.1 of ACI 318-08, UBC 1997, IBC 2006 and NFPA 5000, the required gravity load is 1.2D + f1L where the
seismic force is additive to gravity forces and 0.9D where the seismic force counteracts gravity forces. D is the effect of dead
loads, L is the effect of live loads, and f1 is a factor equal to 0.5 except for garages, areas occupied as places of public
assembly, and all areas where the live load is greater than 100 psf (4.79 kN/m2) where f1 equals 1.0.
C14.2.4.2.5 Relative energy dissipation ratio. This concept is illustrated in Figure C14.2.4.2.2-1 for the third loading cycle
to the limiting drift ratio required by Section 14.2.4.7.4, 14.2.4.7.5 or 14.2.4.7.6, as appropriate.
Figure 14.2.4.2.2-2 illustrates several possible components of the displacement . for a wall that is effectively solid while
Figure C14.2.4.2.2-3 illustrates two possibly undesirable components of the displacement .. Regardless of the mode of
deformation of the wall, the lateral force causes the wall at B to displace horizontally by an amount .. The drift ratio is the
angular rotation of the wall chord with respect to the vertical and for the setup shown equals . / hw where hw is the wall
height and is equal to the distance between the foundation support at A and the load point at B.
Where prestressing steel is used in wall members, the stress fps in the reinforcement at the nominal and the probable lateral
resistance shall be calculated in accordance with Section 18.7 of ACI 318.
C14.2.4.2.3 Global toughness. These provisions describe acceptance criteria for special precast structural walls based on
validation testing. The requirements of ACI 318 Section 21.1.1.8 concerning toughness cover both to the energy dissipation
of the wall system which, for monolithic construction, is affected primarily by local plastic hinging behavior and the
toughness of the prototype structure as a whole. The latter is termed “global toughness” in these provisions and is a condition
that does not apply to the walls alone. That global toughness requirement can be satisfied only though analysis of the
performance of the prototype structure as a whole when the walls perform to the criteria specified in these provisions.
The required gravity load for global toughness evaluations is the value given by these provisions. For conformity with
Section 9.2.1 of ACI 318-08, UBC 1997, IBC 2006 and NFPA 5000, the required gravity load is 1.2D + f1L where the
seismic force is additive to gravity forces and 0.9D where the seismic force counteracts gravity forces. D is the effect of dead
loads, L is the effect of live loads, and f1 is a factor equal to 0.5 except for garages, areas occupied as places of public
assembly, and all areas where the live load is greater than 100 psf (4.79 kN/m2) where f1 equals 1.0.
C14.2.4.2.5 Relative energy dissipation ratio. This concept is illustrated in Figure C14.2.4.2.5 for the third loading cycle
to the limiting drift ratio required by Section 14.2.4.7.4, 14.2.4.7.5 or 14.2.4.7.6, as appropriate. For Figure C14.2.4.2.5, it is
assumed that the test module has exhibited different initial stiffnesses, K and K’, for positive and negative lateral forces and
that the peak lateral resistances for the third cycle for the positive and negative loading directions, E1 and E2, also differ. The
area of the hysteresis loop for the third cycle, Ah, is hatched. The circumscribing figure consists of two parallelograms,
ABCD and DFGA. The slopes of the lines AB and DC are the same as the initial stiffness, K, for positive loading and the
slopes of the lines DF and GA are the same as the initial stiffness, K', for negative loading. The relative energy dissipation
ratio concept is similar to the equivalent damping concept used in Section 17.8.3 of the ASCE/SEI 7-05 for required tests of
seismic isolation systems.
Figure C14.2.4.2.2-1 Showing Definition of drift ratio ..
Figure C14.2.4.2.2-3 Showing Undesirable deformations along horizontal joints: (a) excessive gap opening between panels and (b) shear slip.
Figure C14.2.4.2.2-2 Showing Typical wall deformation components.
Figure C14.2.4.2.2-1 Definition of drift ratio ..
Figure C14.2.4.2.2-2 Typical wall deformation components.
(c) Deformation due to shear
(b) Deformation due to flexure
(d) Deformation due to extension of
reinforcement at foundation to wall
interface
(a) Wall and loading
Figure C14.2.4.2.2-3 Undesirable deformations along horizontal joints:
(a) excessive gap opening between panels and (b) shear slip.
Figure C14.2.4.2.5 Showing Relative energy dissipation ratio.
For a given cycle the relative energy dissipation ratio, ß, is the area, Ah, inside the lateral force-drift ratio loop for the module,
divided by the area of the effective circumscribing parallelograms ABCD and DFGA. The areas of the parallelograms equal
the sum of the absolute values of the lateral force strengths, E1 and E2, at the drift ratios .1 and .2 multiplied by the sum of the
absolute values for the drift ratios .1' and .2'.
C14.2.4.3 Scope and general requirements. While only ACI Committee 318 can determine the requirements necessary
for precast walls to meet the provisions of ACI 318 Section 21.1.1.8, ACI 318 Section 1.4 already permits the building
official to accept wall systems, other than those explicitly covered by ACI 318 Chapter 21, provided specific tests, load
factors, deflection limits, construction procedures and other pertinent requirements have been established for acceptance of
such systems consistent with the intent of the code. The purpose of these provisions is to provide a framework that
establishes the specific tests, load factors, deflection limits and other pertinent requirements appropriate for acceptance, for
regions of high seismic risk or for structures assigned to high seismic performance or design categories, of precast wall
systems, including coupled wall systems, not satisfying all the requirements of ACI 318 Chapter 21. For regions of moderate
seismic risk or for structures assigned to intermediate seismic performance or design categories, less stringent provisions than
those specified here are appropriate.
These provisions assume that the precast wall system to be tested has details differing from those prescribed by ACI 318
Section 21.9 for conventional monolithic reinforced concrete construction. Such walls may, for example, involve the use of
precast elements, precast prestressed elements, post-tensioned reinforcement, or combinations of those elements and
reinforcement.
Figure C14.2.4.2.5 Relative energy dissipation ratio.
For monolithic reinforced concrete walls a fundamental design requirement of ACI 318 Chapter 21 is that walls with hw/lw
exceeding1.0 be proportioned so that their inelastic response is dominated by flexural action on a critical section located near
the base of the wall. That fundamental requirement is retained in these provisions. The reason is that tests on modules, as
envisioned in these provisions, cannot be extrapolated with confidence to the performance of panelized walls of proportions
differing from those tested for the development of ACI 318 Chapter 21 if the shear-slip displacement pattern of Figure
C14.2.4.2.2.3, or the shear deformation response of Figure C14.2.4.2.2.2, governs the response developed in the test on the
module. Two other fundamental requirements of ACI 318 Chapter 21 are for ties around heavily strained boundary element
reinforcement and the provision of minimum amounts of uniformly distributed horizontal and vertical reinforcement in the
web of the wall. Ties around boundary element reinforcement to inhibit its buckling in compression are required where the
strain in the extreme compression fiber is expected to exceed some critical value. Minimum amounts of uniformly distributed
horizontal and vertical reinforcement over the height and length of the wall are required to restrain the opening of inclined
cracks and allow the development of the drift ratios specified in Sections 14.2.4.7.4, 14.2.4.7.5 and 14.2.4.7.6. Deviations
from those tie and distributed reinforcement requirements are possible only if a theory is developed that can substantiate
reasons for such deviations and that theory is tested as part of the validation testing.
C14.2.4.3.1. These provisions are not intended for use with existing construction or for use with walls that are designed to
conform to all the requirements of ACI 318 Section 21.9. The criteria of these provisions are more stringent than those for
walls designed to ACI 318 Section 21.9. Some walls designed to Section 21.9, and having low height to length ratios, may
not meet the drift ratio limits of Equation 14.2.4-1 because their behavior may be governed by shear deformations. The
height to length ratio of 0.5 is the least value for which Equation 14.2.4-1 is applicable.
C14.2.4.3.3 For acceptance, the results of the tests on each module must satisfy the acceptance criteria of Section 14.2.4.9.
In particular, the relative energy dissipation ratio calculated from the measured results for the third cycle between the
specified limiting drift ratios must equal or exceed 1/8. For uncoupled walls, relative energy dissipation ratios increase as the
drift ratio increases. Tests on slender monolithic walls have shown relative energy dissipation ratios, derived from rotations at
the base of the wall, of about 40-45 percent at large drifts. The same result has been reported even where there has been a
significant opening in the web of the wall on the compression side. For 0.020 drift ratios and walls with height to length
ratios of 4, relative energy dissipation ratios have been computed as 30, 18, 12, and 6 percent, for monolithic reinforced
concrete, hybrid reinforced/post-tensioned prestressed concrete with equal flexural strengths provided by the prestressed and
deformed bar reinforcement, hybrid reinforced/post-tensioned prestressed concrete with 25 percent of the flexural strength
provided by deformed bar reinforcement and 75 percent by the prestressed reinforcement, and post-tensioned prestressed
concrete special structural walls, respectively. Thus, for slender precast uncoupled walls of emulative or non-emulative
design it is to be anticipated that at least 35 percent of the flexural capacity at the base of the wall needs to be provided by
deformed bar reinforcement if the requirement of a relative energy dissipation ratio of 1/8 is to be achieved. However, if
more than about 40 percent of the flexural capacity at the base of the wall is provided by deformed bar reinforcement, then
the self-centering capability of the wall following a major event is lost and that is one of the prime advantages gained with
the use of post-tensioning. For squat walls with height to length ratios between 0.35 and 0.69 the relative energy dissipation
has been reported as remaining constant at 23 percent for drifts between that for first diagonal cracking and that for a postpeak
capacity of 80 percent of the peak capacity. Thus, regardless of whether the behavior of a wall is controlled by shear or
flexural deformations a minimum relative energy dissipation ratio of 1/8 is a realistic requirement.
For coupled wall systems, theoretical studies and tests have demonstrated that the 1/8 relative energy dissipation ratio can be
achieved by using central post-tensioning only in the walls and appropriate energy dissipating coupling devices connecting
adjacent vertical wall boundaries.
C14.2.4.3.3.4. The ASCE/SEI 7-05 allowable story drift limits are the basis for the drift limits of IBC 2006 and NFPA 5000.
Allowable story drifts, .a, are specified in Table 1617.3 of IBC 2006 and likely values are discussed in the Commentary to
Section 14.2.4.7.4. The limiting initial drift ratio consistent with .a equals .a/fCdhw, where f is the strength reduction factor
appropriate to the condition, flexure or shear, that controls the design of the test module. For example, for .a/hw equal to
0.015, the required deflection amplification factor Cd of 5, and f equal to 0.9, the limiting initial drift ratio, corresponding to
B in Figure C14.2.4.9.1, is 0.0033. The use of a f value is necessary because the allowable story drifts of the IBC are for the
design seismic load effect, E, while the limiting initial drift ratio is at the nominal strength, En , which must be greater than
E/f. The load-deformation relationship of a wall becomes significantly non-linear before the applied load reaches Ent. While
the load at which that non-linearity becomes marked depends on the structural characteristics of the wall, the response of
most walls remains linear up to about 75 percent of Ent.
C14.2.4.3.3.5. The criteria of Section 14.2.4.9 are for the test module. In contrast, the criterion of Section 14.2.4.3.3.5 is for
the structural system as a whole and can be satisfied only by the philosophy used for the design and analysis of the building
as a whole. The criterion adopted here is similar to that described in the last paragraph of R21.1.1 of ACI 318 and the intent
is that test results and analyses demonstrate that the structure, after cycling three times through both positive and negative
values of the limiting drift ratio specified in Section 14.2.4.7.4, 14.2.4.7.5 or 14.2.4.7.6, as appropriate, is still capable of
supporting the gravity load specified as acting on it during the earthquake.
Figure C14.2.4.9.1 Quantities used in evaluating acceptance criteria.
Figure C14.2.4.9.1 Quantities used in evaluating acceptance criteria.
C14.2.4.4 Design Procedure.
C14.2.4.4.1. The test program specified in these provisions is intended to verify an existing design procedure for precast
structural walls for a specific structure or for prequalifying a generic type of special precast wall system for construction in
general. The test program is not for the purpose of creating basic information on the strength and deformation properties of
such systems for design purposes. Thus, the test modules should not fail during the validation testing, a result that is the
opposite of what is usually necessary during testing in the development phase for a new or revised design procedure. For a
generic precast wall system to be accepted based on these provisions, a rational design procedure is to have been developed
prior to this validation testing. The design procedure is to be based on a rational consideration of material properties and
force transfer mechanisms, and its development will usually require preliminary and possibly extensive physical testing that
is not part of the validation testing. Because special wall systems are likely to respond inelastically during design-level
ground shaking, the design procedure must consider wall configuration, equilibrium of forces, compatibility of deformations,
the magnitudes of the lateral drifts, reversed cyclic displacements, the relative values of each limiting engineering design
criteria (shear, flexure and axial load) and use appropriate constitutive laws for materials that include considerations of
effects of cracking, loading reversals and inelasticity.
The effective initial stiffness of the structural walls is important for calculating the fundamental period of the prototype
structure. The procedure used to determine the effective initial stiffness of the walls is to be verified from the validation test
results as described in Section 14.2.4.7.11.
Provisions Sections 14.2.4.4.1.1 through 14.2.4.4.1.3 state the minimum procedures to be specified in the design procedure
prior to the start of testing. The Authority Having Jurisdiction may require that more details be provided in the design
procedure than those of Sections 14.2.4.4.1.1 through 14.2.4.4.1.3 prior to the start of testing.
C14.2.4.4.2. The justification for the small number of test modules, specified in Section 14.2.4.5.1 is that a previously
developed rational design procedure is being validated by the test results. Thus, the test modules for the experimental
program must be designed using the procedure intended for the prototype wall system and strengths must be predicted for the
test modules before the validation testing is started.
Figure C14.2.4.5.1 (a) Showing Coupled wall test module with coupling beams and (b) Showing Coupled wall test module with vertical mechanical couplers.
C14.2.4.5 Test Modules.
C14.2.4.5.1. One module must be tested for each limiting engineering design criterion, such as shear, or axial load and
flexure, for each characteristic configuration of walls. Thus, in accordance with Section 14.2.4.4.3 if the test on the module
results in a maximum shear stress of 3vfc’ then the maximum shear stress that can be used in the prototype is that same value.
Each characteristic in-plane configuration of walls, or coupled walls, in the prototype structure must also be tested. Thus, as
a minimum for one-way structural walls, two modules with the configuration shown in Figure C14.2.4.2.2-1, and, for one
way coupled walls, two modules with the configuration shown in either Figure C14.2.4.5.1(a) or in Figure C14.2.4.5.1(b),
must be tested. In addition, if intersecting wall systems are to be used then the response of the wall systems for the two
orthogonal directions needs to be tested. For two-way wall systems and coupled wall-frame systems, testing of configurations
other than those shown in Figures C14.2.4.2.2-1 and C14.2.4.5.1 may be appropriate when it is difficult to realistically model
the likely dominant earthquake deformations using orthogonal direction testing only.
This provision should not be interpreted as implying that only two tests will need to be made to qualify a generic system.
During the development of that system it is likely that several more tests will have been made, resulting in progressive
refinements of the mathematical model used to describe the likely performance of the generic structural wall system and its
construction details. Consequently, only one test of each module type for each limiting engineering design condition, at a
specified minimum scale and subjected to specific loading actions, may be required to validate the system. Further, as stated
in Section 14.2.4.9.1, if any one of those modules for the generic wall system fails to pass the validation testing required by
these provisions, then the generic wall system has failed the validation testing
In most prototype structures, a slab is usually attached to the wall and, as demonstrated by the results of the PRESSS building
test, the manner in which the slab is connected to the wall needs to be carefully considered. The connection needs to be
adequate to allow the development of story drifts equal to those anticipated in these provisions. However, in conformity with
common practice for the sub-assemblage tests used to develop the provisions of Chapter 21of ACI 318, there is no
requirement for a slab to be attached to the wall of the test module. The effect of the presence of the slab should be examined
in the development program that precedes the validation testing.
C14.2.4.5.3. Test modules need not be as large as the corresponding walls in the prototype structure. The scale of the test
modules, however, must be large enough to capture all the complexities associated with the materials of the prototype wall,
its geometry and reinforcing details, load transfer mechanisms, and joint locations. For modules involving the use of precast
elements, for example, scale effects for load transfer through mechanical connections should be of particular concern. The
issue of the scale necessary to capture fully the effects of details on the behavior of the prototype should be examined in the
development program that precedes the validation testing.
drift
angle, .
angle, .
drift
angle, .
drift
centrally post-tensioned
coupling beams
grouted deformed top bars
deflected
configuration
central unbonded
post-tensioning
central unbonded
post-tensioning
undeflected
position
relative vertical
deflection
drift
drift
undeflected
position
mechanical coupling
devices
deflected
configuration
Figure C14.2.4.5.1 (a) Coupled wall test module with coupling beams;
(b) Coupled wall test module with vertical mechanical couplers.
C14.2.4.5.4. It is to be expected that for a given generic precast wall structure, such as an unbonded centrally post-tensioned
wall constructed using multiple precast or precast pretensioned concrete wall panels, validation testing programs will initially
use specific values for the specified strength of the concrete and reinforcement in the walls, the layout of the connections
between panels, the location of the post-tensioning, the location of the panel joints, and the design stresses in the wall.
Pending the development of an industry standard for the design of such walls, similar to the standard for special hybrid
moment frames, specified concrete strengths, connection layouts, post-tensioning amounts and locations, etc., used for such
walls will need to be limited to the values and layouts used in the validation testing programs.
C14.2.4.5.5. For walls constructed using precast or precast/prestressed panels and designed using non-emulative methods,
the response under lateral load can change significantly with joint opening (Figure C14.2.4.2.2-2d and Figure C14.2.4.2.2-
3a). The number of panels used to construct a wall depends on wall height and design philosophy. If, in the prototype
structure, there is a possibility of horizontal joint opening under lateral loading at a location other than the base of the wall,
then the consequences of that possibility need to be considered in the development and validation test programs. Joint
opening at locations other than the base can be prevented through the use of capacity design procedures.
C14.2.4.5.6. The significance of the magnitude of the gravity load that acts simultaneously with the lateral load needs to be
addressed during the validation testing if the development program suggests that effect is significant.
C14.2.4.5.7. Details of the connection of walls to the foundation are critical, particularly for non-emulative wall designs.
The deformations that occur at the base of the wall due to plastic hinging or extension of the reinforcing bars or posttensioning
steel crossing the wall to foundation interface, (Figure C14.2.4.2.2-2d), are in part determined by details of the
anchorage and the bonding of those reinforcements on either side of the interface. Grout will be normally used to bed panels
on the foundation and the characteristics of that grout in terms of materials, strength and thickness, can have a large effect on
wall performance. The typical grout pad with a thickness of 1 inch (25 mm) or less can be expected to provide a coefficient
of friction of about 0.6 under reversed loadings. Pads with greater thickness and without fiber reinforcement exhibit lesser
coefficients of friction. Adequate frictional resistance is essential to preventing undesirable shear-slip deformations of the
type shown in Figure C14.2.4.2.2.3(b).
C14.2.4.5.8. The geometry of the foundations need not duplicate that used in the prototype structure. However, the
geometric characteristics of the foundations (width, depth and length) need to be large enough that they do not influence the
behavior of the test module.
C14.2.4.6 Testing Agency. In accordance with the spirit of the requirements of Sections 1.3.5 and 1.4 of ACI 318, it is
important that testing be carried out by a recognized independent testing agency, approved by the agency having jurisdiction
and that the testing and reporting be supervised by a registered design professional familiar with the proposed design
procedure and experienced in testing and seismic structural design.
C14.2.4.7 Test Method. The test sequence is expressed in terms of drift ratio, and the initial ratio is related to the likely
range of linear elastic response for the module. That approach, rather than testing at specific drift ratios of 0.005, 0.010, etc.,
is specified because, for modules involving prestressed concrete, the likely range of elastic behavior varies with the prestress
level.
An example of the test sequence specified in Sections 14.2.4.7.2 through 14.2.4.7.6 is illustrated in Figure C14.2.4.7. The
sequence is intended to ensure that displacements are increased gradually in steps that are neither too large nor too small. If
steps are too large, the drift capacity of the system may not be determined with sufficient accuracy.
If the steps are too small, the system may be unrealistically softened by loading repetitions, resulting in artificially low
maximum lateral resistances and artificially high maximum drifts. Also, when steps are too small, the rate of change of
energy stored in the system may be too small compared with the change occurring during a major event. Results, using such
small steps, can mask undesirable brittle failure modes that might occur in the inelastic response range during a major event.
Because significant diagonal cracking is to be expected in the inelastic range in the web of walls, and in particular in squat
walls, the pattern of increasing drifts used in the test sequence can markedly affect diagonal crack response in the post-peak
range of behavior.
The drift capacity of a building in a major event is not a single quantity, but depends on how that event shakes the structure.
In the forward near field, a single pulse may determine the maximum drift demand, in which case a single large drift demand
cycle for the test module would give the best estimation of the drift capacity. More often, however, many small cycles
precede the main shock and that is the scenario represented by the specified loading.
Figure C14.2.4.7 Showing Example of specified test sequence.
There is no requirement for an axial load to be applied to the wall simultaneously with the application of the lateral
displacements. In many cases it will be conservative not to apply axial load because, in general, the shear capacity of the
wall and the resistance to slip at the base of the wall increase as the axial load on the wall increases. However, as the height
of the wall increases and the limiting drift utilized in the design of the wall increases, the likelihood of extreme fiber crushing
in compression at maximum drift increases, and the importance of the level of axial load increases. The significance of the
level of axial loading should be examined during the development phase.
C14.2.4.7.4 For the response of a structure to the design seismic shear force, building codes (e.g., UBC 97, IBC 2006 or
NFPA 5000) or recommended provisions (e.g., ASCE/SEI 7-05 and FEMA 356) specify a maximum allowable drift.
However, structures designed to meet that drift limit may experience greater drifts under the design basis earthquake ground
motion and are likely to experience greater drifts under the risk-targeted maximum considered earthquake ground motion. In
addition to the characteristics of the ground motion, actual drifts will depend on the strength of the structure, its initial elastic
stiffness, and the ductility expected for the given lateral load resisting system. Specification of suitable limiting drifts for the
test modules requires interpretation and allowance for uncertainties in the assumed ground motions and structural properties.
In IBC 2006, the design seismic shear force applied at the base of a building is related directly to its weight and the design
elastic response acceleration, and inversely to a response modification factor, R. That R factor increases with the expected
ductility of the lateral force resisting system of the building. Special structural walls satisfying the requirements of Sections
21.1 and 21.9 are assigned an R value of 6 when used in a building frame system and a value of 5 when used in a bearing
wall system. They are also assigned allowable story drift ratios that are dependent on the hazard to which the building is
exposed. When the design seismic shear force is applied to a building, the building responds inelastically and the resultant
computed drifts, (the design story drifts), must be less than a specified allowable drift. Additional guidance is given in
FEMA 356 where the deformations for rectangular walls with height to length ratios greater than 2.5, and flanged wall
sections with height to length ratios greater than 3.5, are to be assumed to be controlled by flexural actions. When structural
walls are part of a building representing a substantial hazard to human life in the event of a failure, the allowable story drift
ratio for shear controlled walls is 0.0075 and for flexure controlled walls is a function of the plastic hinge rotation at the base
of the wall. For flexure controlled walls values range up to a maximum of about 0.02 for walls with confined boundary
elements with low reinforcement ratios and shear stress less than 3vfc’.
Figure C14.2.4.7 Example of specified test sequence.
To compensate for the use of the R value, IBC Section 1617.4.6 requires that the drift determined by an elastic analysis for
the code-prescribed seismic forces be multiplied by a deflection amplification factor, Cd ,to determine the design story drift
and that the design story drift must be less than the allowable story drift. In building frame systems, structural walls
satisfying the requirements of Section 21.9 of ACI 318 are assigned a Cd value of 5. However, research has found that design
story drift ratios determined in the foregoing manner may be too low. Drift ratios of 6 times IBC-calculated values, (rather
than 5), are more representative of the upper bounds to expected drift ratios. The value of 6 is also in agreement with the
finding that the drift ratio of an inelastic structure is approximately the same as that of an elastic structure with the same
initial period. For flexure controlled walls the value of 6/5 times the present IBC limits on calculated drift ratio, would lead
to a limit on real drift ratios of up to 0.024.
Duffy et al. reviewed experimental data for shear walls to define post-peak behavior and limiting drift ratios for walls with
height to length ratios between 0.25 and 3.5. Seo et al. re-analyzed the data of Duffy et al. together with data from tests
conducted subsequent to the analysis of Duffy et al. Duffy et al. established that for squat walls with web reinforcement
satisfying ACI 318-02 requirements and height to length ratios between 0.25 and 1.1, there was a significant range of
behavior for which drifts were still reliable in the post-peak response region. Typically the post-peak drift increased by 0.005
for a 20 percent degradation in capacity under cyclic loading. For greater values of degradation, drifts were less reliable.
That finding has also been confirmed through tests conducted by Hidalgo et al. (2002) on squat walls with effective height to
length ratios ranging between 0.35 and 1.0. Values of the drift ratio of the walls at inclined cracking and at peak capacity
varied little with web reinforcement. By contrast, drifts in the post-peak range were reliable to a capacity equal to 80 percent
of the peak capacity and were 0.005 greater than the drifts at peak capacity provided the walls contained horizontal and
vertical web reinforcement equal to 0.25 percent.
From an analysis of the available test data, and from theoretical considerations for a wall rotating flexurally about a plastic
hinge at its base, Seo et al. concluded that the limiting drift at peak capacity increased almost linearly with the height to
length ratio of the wall. When the additional post peak drift capacity for walls with adequate web reinforcement was added
to the drift at peak capacity, the total available drift capacity in percent was given by 1.0 = 0.67 (hw / lw) + 0.5 = 3.0 where hw
is the height of the wall, and lw is the length of the wall.
The data from the tests of Hidalgo et al. (2002) suggest that while that formula is correct for squat walls, the lower limit on
drift can be decreased to 0.8 as specified in these provisions and that the use of that formula should be limited to walls with
height to length ratios equal to or greater than 0.5. For wall height to length ratios less than 0.5, the behavior is controlled
principally by shear deformations (Figure C14.2.4.2.2.2c), and Equation 14.2.4-1 should not be used. The upper value of
0.030 for the drift ratio was somewhat optimistic because the data were for walls with height to length ratios equal to or less
than 3.5 and subsequent tests have shown that the upper limit of 2.5, as specified in Equation 14.2.4.1, is a more realistic
limit.
C14.2.4.7.5 The design capacity for coupled wall systems must be developed by the drift ratio corresponding to that for the
wall with the least hw/lw value. However, it is desirable that testing be continued to the drift given by Equation 14.2.4-1 for
the wall with the greatest hw/lw in order to assess the reserve capacity of the coupled wall system.
C14.2.4.7.6 The drift limits of Equation 14.2.4.1 are representative of the maximum that can be achieved by walls designed
to ACI 318. The use of smaller drift limits is appropriate if the designer wishes to use performance measures less than the
maximum permitted by ACI 318. Examples are the use of reduced shear stresses so that the likelihood of diagonal cracking
of the wall is minimized or reduced compressive stresses in the boundary elements of the wall so that the risk of crushing is
reduced. Nonlinear time history analyses for the response to a suite of risk-targeted maximum considered earthquakes
(MCER) ground motions, rather than 1.5 times a suite of the corresponding design basis earthquake (DBE) ground motions, is
required because the drifts for the response to the MCER motion can be significantly larger than 1.5 times the drifts for the
response to the DBE motions.
C14.2.4.7.10 In many cases, data additional to the minimum specified in Section 14.2.4.7.7 may be useful to confirm both
design assumptions and satisfactory response. Such data include relative displacements, rotations, curvatures, and strains.
C14.2.4.8 Test Report. The test report must be sufficiently complete and self-contained for a qualified expert to be satisfied
that the tests have been designed and carried out in accordance with these criteria, and that the results satisfy the intent of
these provisions. Sections 14.2.4.8.1.1 through 14.2.4.8.1.11 state the minimum evidence to be contained within the test
report. The authority having jurisdiction or the registered design professional supervising the testing may require that
additional test information be reported.
C14.2.4.9 Test Module Acceptance Criteria.
The requirements of this clause apply to each module of the test program and not to an average of the results of the program.
Figure C14.2.4.9.1 illustrates the intent of this clause.
Figure C14.2.4.9.1 Showing Unacceptable hysteretic behavior.
C14.2.4.9.1.1 Where nominal strengths for opposite loading directions differ, as is likely for C-, L- or T- shaped walls, the
criterion of Section 14.2.4.9.1.1 applies separately to each direction.
C14.2.4.9.1.2 At high cyclic-drift ratios, strength degradation is inevitable. To limit the level of degradation so that drift ratio
demands do not exceed anticipated levels, a maximum strength degradation of 0.20Emax is specified. Where strengths differ
for opposite loading directions, this requirement applies independently to each direction.
C14.2.4.9.1.3. If the relative energy dissipation ratio is less than 1/8, there may be inadequate damping for the building as a
whole. Oscillations may continue for some time after an earthquake, producing low-cycle fatigue effects, and displacements
may become excessive.
If the stiffness becomes too small around zero drift ratio, the structure will be prone to large displacements for small lateral
force changes following a major earthquake. A hysteresis loop for the third cycle between peak drift ratios of 1/10 times the
limiting drift ratio given by Equation 14.2.4-1, that has the form shown in Figure C14.2.4.9.1, is acceptable. At zero drift
ratio, the stiffnesses for positive and negative loading are about 11 percent of the initial stiffnesses. Those values satisfy
Section 14.2.4.9.1. An unacceptable hysteresis loop form would be that shown in Figure C14.2.4.9.1 where the stiffness
around zero drift ratio is unacceptably small for both positive and negative loading.
COMMENTARY TO SECTION 14.4.5
C14.4.5 Modifications to Chapter 1 of ACI 530/ASCE 5/TMS 402. The seismic design factors, SDC limits, and height
restrictions of these provisions are based on a combination of testing, analysis, underlying consensus standards, experience,
and consistency with comparable structural systems.
The testing and analysis, described in Tanner et al. (2005a and b) and Varela et al. (2005b), began in 1999 and were
developed as part of an integrated research strategy. This strategy, presented at ICC-ES hearings in 2003 and affirmed in its
essence using performance-based methods similar to those in the 90-percent-complete draft of FEMA P-695 (Applied
Technology Council, 2008), had as its objective the development of seismic design factors consistent with at most a 10
percent probability of collapse under what was essentially equivalent to the maximum considered earthquake ground motion.
That research developed factors of R and Cd equal to 3 with no restrictions on SDC or height. Additional information on that
research is presented in American Society of Testing and Materials (2007), Masonry Standards Joint Committee (2005a and b
and 2008a and b), The Masonry Society (2007), Tanner et al. (2005a and b), and Varela et al. (2006).
Figure C14.2.4.9.1 Unacceptable hysteretic behavior.
Following the initial presentation of this strategy and its associated proposals in the ICC-ES forum, it was discussed
extensively with the BSSC’s Provisions Update Committee and other interested parties including the BSSC’s Code Resource
Support Committee. Those discussions led to a modification of the proposal to R and Cd factors equal to 2, to SDC from A to
C, and to height restrictions of 35 ft for SDC C. These values and their associated restrictions are consistent with a
probability of failure much lower than 10 percent under what was essentially equivalent to the risk-targeted maximum
considered earthquake ground motion (MCER).
Structures of autoclaved aerated concrete (AAC) masonry are designed and constructed using U.S. consensus standards
including material standards (American Society of Testing and Materials, 2007), design provisions, and mandatory
construction requirements (Masonry Standards Joint Committee, 2005a and b and 2008a and b). These U.S. consensus
standards are augmented by refereed documents (The Masonry Society, 2007) and the online recommendations of the
Autoclaved Aerated Concrete Products Association (http://www.aacpa.org/).
In the United States, AAC masonry buildings built with local approvals, under design rules consistent with the consensus
standards, and with heights greater than those permitted by these provisions, have successfully resisted hurricane winds with
no damage.
The seismic design factors, SDC limits, and height restrictions of these provisions are consistent (or even more conservative)
than those assigned to Ordinary Reinforced Masonry Shear Walls of clay or concrete masonry.
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Modifications to Chapter 15, Seismic Design Requirements for
Nonbuilding Structures
TABLE 15.4-2, SEISMIC COEFFICIENTS FOR
NONBUILDING STRUCTURES NOT SIMILAR TO BUILDINGS
Revise the following items as indicated (deletions in strikeout and additions underlined):
Cast-in-place concrete silos, stacks, and
chimneys having walls continuous to the
foundation
15.6.2
3
1.75
3
NL
NL
NL
NL
NL
All other reinforced masonry structures
not similar to buildings
14.4.1
3
2
2.5
NL
NL
NL
50
50
All other nonreinforced masonry
structures not similar to buildings
14.4.1
1.25
2
1.5
NL
NL
50
50
50
Concrete chimneys and stacks
15.6.2
2
1.5
1.5
NL
NL
NL
NL
NL
All other steel and reinforced concrete
distributed mass cantilever structures not
covered herein including stacks,
chimneys, silos, and skirt-supported
vertical vessels that are not similar to
buildings
15.6.2 15.7.10 and
15.7.10.5 a and b.
3
2
2.5
NL
NL
NL
NL
NL
SECTION 15.5.3, STEEL STORAGE RACKS
Replace with the following:
15.5.3 Steel Storage Racks. Steel storage racks supported at or below grade shall be designed in accordance with
Section 2.7 of the ANSI/RMI MH 16.1 standard and its force and displacement requirements.
For storage racks supported above grade, the value of V in Section 2.7.2 of ANSI/RMI MH 16.1 shall not be taken less
than the value of Fp determined in accordance with Section 13.3.1 of this standard, where Rp is taken equal to R, and ap is
taken equal to 2.5.
Alternatively, in addition to the requirements of Section 15.5.1, steel storage racks shall be designed in accordance with
the requirements of Sections 15.5.3.1 through 15.5.3.4
[Sections 15.5.3.1 through 15.5.3.4 are unchanged.]
SECTION 15.6.2, STACKS AND CHIMNEYS
Replace with the following:
15.6.2 Stacks and Chimneys. Stacks and chimneys are permitted to be either lined or unlined and shall be constructed
from concrete, steel, or masonry. Steel stacks, concrete stacks, steel chimneys, concrete chimneys, and liners shall be
designed to resist seismic lateral forces determined from a substantiated analysis using reference documents. Interaction
of the stack or chimney with the liners shall be considered. A minimum separation shall be provided between the liner
and chimney equal to Cd times the calculated differential lateral drift.
For concrete chimneys assigned to Seismic Design Category D, E or F, splices for vertical rebar shall be staggered such
that no more than 50 percent of the bars are spliced at any elevation. Design and detailing of cross-sections in the
regions of breach openings, where the loss of cross-sectional area is greater than 10 percent, shall be performed in one of
the following ways:
a. For vertical force, shear force, and bending moment demands along the vertical direction, design the affected crosssection
using the overstrength factor of 1.5. The following detailing requirements shall be satisfied:
i. The region of such overstrength shall extend above and below (except if the opening is at the base) the
opening(s) by a distance equal to half of the width of the largest opening in the affected region.
ii. Appropriate reinforcement development lengths shall be provided beyond the required region of overstrength.
iii. The jamb regions around each opening shall be detailed using the column tie requirements in Section 7.10.5 of
ACI 318. Such detailing shall extend for a jamb width of a minimum of two times the wall thickness and for a
height of the opening height plus twice the wall thickness above and below the opening, but no less than the
development length of the longitudinal bars. The percentage of longitudinal reinforcement in jamb regions shall
meet the requirements of Section 10.9 of ACI 318 for compression members.
b. Provided that the cross-sectional moment of inertia in the opening region is at least 70 percent of the same above
and below it, it shall be permitted to treat the breach opening region as follows:
i. All detailing requirements listed in Item a. above for the overstrength option shall be followed, in addition to the
ones listed below.
ii. Hoop ties in jamb regions shall be detailed as columns of intermediate moment frames using the requirements
in Section 21.3.5 of ACI 318. The dimensions for jamb region shall be the same as that required in Item a.
above for the overstrength option.
iii. No construction joints within the opening region plus two times the wall thickness above and below the
opening.
iv. Ratio of outer diameter to wall thickness shall not exceed 20 within the opening region.
SECTION 15.7.6, GROUND-SUPPORTED STORAGE TANKS FOR LIQUIDS
Add the following exception to the end of Section 15.7.6.1, General:
EXCEPTION: For Tc > 4 seconds, Sac may be determined by a site-specific study using one or more of the following
methods: (a) the procedures found in Chapter 21, provided such procedures, which rely on ground-motion attenuation
equations for computing response spectra, cover the natural period band containing Tc , (b) ground-motion simulation
methods employing seismological models of fault rupture and wave propagation, and (c) analysis of representative
strong-motion accelerogram data with reliable long-period content extending to periods greater than Tc . However, in no
case shall the value of Sac be taken as less than the minimum of:
1. The value determined in accordance with Equation 15.7-11 using 50 percent of the mapped value of TL from Figure
22-7 or
2. 0.8 times the value determined in accordance with Equation 15.7-11 using the mapped value of TL from Figure 22-7.
In determining the value of Sac, the value of TL shall not be less than 4 seconds.
Commentary to Chapter 15 Modifications
COMMENTARY TO SECTION 15.5.3
C15.5.3 Steel Storage Racks. The two approaches to the design of steel storage racks set forth by the standard are intended
to produce comparable results. The specific revisions to the RMI specification cited in earlier editions of the Provisions and
the detailed requirements of the new ANSI/RMI standard reflect the recommendations of FEMA 460, Seismic Considerations
for Steel Storage Racks Located in Areas Accessible to the Public.
COMMENTARY TO SECTION 15.6.2
C15.6.2 Stacks and Chimneys. The design of stacks and chimneys to resist natural hazards generally is governed by wind
design considerations. The exceptions to this general rule involve locations with high seismicity, stacks and chimneys with
large elevated masses, and stacks and chimneys with unusual geometries. It is prudent to evaluate the effect of seismic loads
in all but those areas with the lowest seismicity. Although not specifically required, it is recommended that the special
seismic details required elsewhere in the standard be considered for application to stacks and chimneys.
Concrete chimneys have low ductility, and their seismic behavior is especially critical in the opening regions due to inherent
reduction in strength and loss of confinement for vertical reinforcement in the jamb regions around the openings. Spectacular
earthquake-induced chimney failures have occurred in recent history (in Turkey in 1999) and have been attributed to
strength/detailing problems (Kilic and Sozen, 2003). Therefore, the R value of 3 traditionally used in ASCE/SEI 7-05 for
concrete stacks and chimneys is reduced to 2 and detailing requirements for breach openings are added in the 2009 NEHRP
Recommended Seismic Provisions.
Guyed steel stacks and chimneys are generally lightweight. As a result, the design loads due to natural hazards generally are
governed by wind. On occasion, large flares or other elevated masses located near the top may require in-depth seismic
analysis. Although it does not specifically address seismic loading, Chapter 6 of Troitsky (1982) provides a methodology
appropriate for resolution of the seismic forces defined in the standard.
COMMENTARY TO SECTION 15.7.6.1
C15.7.6.1 General. The response of ground storage tanks to earthquakes is well documented by Housner, Mitchell and
Wozniak, Veletsos, and others. Unlike building structures, the structural response of these tanks is influenced strongly by the
fluid-structure interaction. Fluid-structure interaction forces are categorized as sloshing (convective) and rigid (impulsive)
forces. The proportion of these forces depends on the geometry (height-to-diameter ratio) of the tank. API 650, API 620,
AWWA D100, AWWA D110, AWWA D115, and ACI 350.3 provide the data necessary to determine the relative masses
and moments for each of these contributions.
The standard requires that these structures be designed in accordance with the prevailing reference documents, except that the
height of the sloshing wave, ds
, must be calculated using Equations 15.7-13. Note that API 650 and AWWA D100 include
this requirement in their latest editions.
Equations 15.7-10 and 15.7-11 provide the spectral acceleration of the sloshing liquid for the constant-velocity and constantdisplacement
regions of the response spectrum, respectively. The 1.5 factor in these equations is an adjustment for
0.5 percent damping. An exception in the use of Equation 15.7-11 was added for the 2009 NEHRP Recommended Seismic
Provisions. Actual site-specific studies carried out since the introduction of the TL requirements of ASCE/SEI 7-05 indicate
that the mapped values of TL are extremely conservative. Because a revision of the TL maps is a time-consuming task that
would not be possible during the 2009 Provisions update cycle, an exception was added to allow the use of site-specific
values that are less than the mapped values with a floor of 4 seconds or one-half the mapped value of TL. The exception was
added under Section 15.7.6 because TL is a tank issue. Discussion of the site-specific procedures can be found in the Part 2
Commentary for Chapter 22.
ADDITIONAL REFERENCE FOR CHAPTER 15 COMMENTARY
Kilic, S., and M. Sozen. 2003. “Evaluation of Effect of August 17, 1999, Marmara Earthquake on Two Tall Reinforced
Concrete Chimneys,” ACI Structural Journal, 100(3).
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Modification to Chapter 16,
Seismic Response History Procedures
SECTION 16.1.3.2, THREE-DIMENSIONAL ANALYSIS
Replace with the following:
16.1.3.2 Three-Dimensional Analysis. Where three-dimensional analyses are performed, ground motions shall consist
of pairs of appropriate horizontal ground motion acceleration components that shall be selected and scaled from
individual recorded events. Appropriate ground motions shall be selected from events having magnitudes, fault
distances, and source mechanisms that are consistent with those that control the risk-targeted maximum considered
earthquake (MCER). Where the required number of recorded ground motion pairs is not available, appropriate simulated
ground motion pairs are permitted to be used to make up the total number required. For each pair of horizontal ground
motion components, a square root of the sum of squares (SRSS) spectrum shall be constructed by taking the SRSS of the
5-percent-damped response spectra for the scaled components (for direct scaling, an identical scale factor is applied to
both components of a pair). Each pair of motions shall be scaled such that for each period between 0.2T and 1.5T, the
average of the SRSS spectra from all horizontal component pairs does not fall below the corresponding ordinate of the
MCER response spectrum determined in accordance with Section 11.4.5 or 11.4.7.
At sites within 5 km of an active fault that controls the hazard, each pair of components shall be rotated to the faultnormal
and fault-parallel direction of the causative fault and shall be scaled so that the average of the fault-normal
components is not less than the MCER response spectrum for each period between 0.2T and 1.5T.
Commentary to Chapter 16 Modification
COMMENTARY TO SECTION 16.1.3.2
C16.1.3.2 Three-dimensional Analyses. One key change to the ground motion design requirements developed by the
BSSC’s Seismic Design Procedure Review Group (SDPRG) for the 2009 NEHRP Recommended Seismic Provisions is the
use of maximum direction ground motions. In addition to changing the design values defined in Chapter 11 and used
throughout the Provisions, implementing maximum direction ground motions affects the previous ground motion scaling
rules specified in Section 16.1.3.2. Studies (Maffei and Hashemi, 2008) of 50 ground motions of M6.5-M7.9 earthquakes for
both far-field and near-field records and for periods in the range of 0.1 to 3.0 seconds indicate that the maximum direction of
ground motion is slightly less than the SRSS of the two components with the SRSS spectrum tending to be approximately
1.16 times the maximum direction spectrum.
For each of the 50 ground motions, the maximum response of a single-degree-of-freedom (SDOF) oscillator (assuming 5-
percent damping) was determined for ground motion orientations from 0 to 90 degrees (in one-degree increments) and was
compared to the associated SRSS of maximum response. The ratios of the SRSS of maximum response and the maximum
amplitude of the response for varying parameters are given in Tables C16.1.3.2-1 through C16.1.3.2-3.
Table C16.1.3.2-1 Ratio of SRSS of Maximum Response to Maximum
Amplitude as a Function of SDOF Period
SDoF Period
Number of Data
Points
Ratio-Mean
Ratio –Standard
Deviation
0.1 sec
50
1.19
0.077
0.3 sec
50
1.16
0.068
1.0 sec
50
1.14
0.067
3.0 sec
50
1.13
0.077
Average
200
1.16
0.076
Table C16.1.3.2-2 Ratio of SRSS of Maximum Response to Maximum
Amplitude as a Function of Ground Motion Records
Ground Motion
Number of Data
Points
Ratio-Mean
Ratio -Standard
Deviation
Far-Field
88
1.16
0.067
Near-Field
112
1.15
0.078
Average
200
1.16
0.076
Table C16.1.3.2-3 Ratio of SRSS of Maximum Response to Maximum
Amplitude as a Function of Site Class
Site Class
Number of Data
Points
Ratio-Mean
Ratio -Standard
Deviation
B
8
1.15
0.066
C
84
1.15
0.072
D
108
1.16
0.073
Average
200
1.16
0.076
The modified scaling requirements simplify phrasing of existing language by replacing 10 percent less than 1.16 times the
MCER response spectrum with the MCER response spectrum, itself, resulting in an effective “1.0” multiplier. This effective
multiplier comes from (0.9)(1.16) ˜ 1.0.
However, for sites within approximately 5 km of an active fault that controls the ground-motion hazard, the near field strongmotion
database indicates that the fault-normal (FN) direction is (or is close to) the direction of maximum ground motion for
periods around 1.0 second and greater (Huang et al., 2008; Watson-Lamprey and Boore, 2007). In this case, the two
horizontal components of a selected record are to be transformed so that one component is the motion in the FN direction and
the other component is the motion in the fault-parallel (FP) direction. Scaling so that the average FN component response
spectrum is at the level of the MCER response spectrum ensures that the FN components will not be underestimated, which
would happen if the SRSS rule was applied at short distances. The same scale factor selected for the FN component of a
given record is used for the FP component also.
ADDITIONAL REFERENCES FOR CHAPTER 16 COMMENTARY
Huang, Y. N., A. Whittaker, and N. Luco. 2007. “NGA Relationships, USGS Seismic Hazard Maps, Near-Fault Ground
Motions and Site Effects: BSSC Project 07 Final Draft Report. BSSC, Washington, D.C.
Maffei, J., and A. Hashemi. 2008. Personal Communication.
Watson-Lamprey, J. A., and D. M. Boore. 2007. “Beyond SaGMRotI: Conversion to SaArb, SaSN, and SaMaxRot,” Bulletin of the
Seismology Society of America, 97:1511-1524.
Modifications to Chapter 18, Seismic Design Requirements for Structures
with Damping Systems
SECTION 18.3.1, NONLINEAR RESPONSE HISTORY PROCEDURE
Replace with the following:
18.3.1 Nonlinear Response History Procedure. A nonlinear response history (time history) analysis shall utilize a
mathematical model of the structure and the damping system as provided in Chapter 16 and this section. The model
shall directly account for the nonlinear hysteretic behavior of elements of the structure and the damping devices to
determine its response, through methods of numerical integration, to suites of ground motions compatible with the
design response spectrum for the site.
The analysis shall be performed in accordance with Chapter 16 together with the requirements of this section. Inherent
damping of the structure shall not be taken greater than 5 percent of critical unless test data consistent with levels of
deformation at or just below the effective yield displacement of the seismic-force-resisting system support higher values.
If the calculated force in an element of the seismic force-resisting system does not exceed 1.5 times its nominal strength,
that element is permitted to be modeled as linear.
18.3.1.1 Damping Device Modeling. Mathematical models of displacement-dependent damping devices shall include
the hysteretic behavior of the devices consistent with test data and accounting for all significant changes in strength,
stiffness, and hysteretic loop shape. Mathematical models of velocity-dependent damping devices shall include the
velocity coefficient consistent with test data. If this coefficient changes with time and/or temperature, such behavior
shall be modeled explicitly. The elements of damping devices connecting damper units to the structure shall be included
in the model.
Exception: If the properties of the damping devices are expected to change during the duration of the response
history analysis, the dynamic response is permitted to be enveloped by the upper and lower limits of device
properties. All these limit cases for variable device properties must satisfy the same conditions as if the time
dependent behavior of the devices were explicitly modeled.
18.3.1.2 Response Parameters. For each ground motion analyzed, individual response parameters consisting of the
maximum value of the individual member forces, member inelastic deformations and story drifts at each story shall
be determined. Moreover, for each ground motion used for response history analysis, individual response
parameters consisting of the maximum value of the discrete damping device forces, displacements, and velocities, in
the case of velocity-dependent devices, shall be determined.
If at least seven ground motions are used for response history analysis, the design values of the damping device forces,
displacements, and velocities are permitted to be taken as the average of the values determined by the analyses. If fewer
than seven ground motions are used for response history analysis, the design damping device forces, displacements and
velocities shall be taken as the maximum value determined by the analyses. A minimum of three ground motions shall
be used.
SECTION 18.3.2, NONLINEAR STATIC PROCEDURE
Replace with the following:
18.3.2 Nonlinear Static Procedure. Nonlinear static procedures may be used to construct the lateral forcedisplacement
curve of the seismic-force-resisting system in lieu of the elastoplastic curve assumed in the response
spectrum procedure and in the equivalent lateral force procedure. When nonlinear static procedures is used, the
nonlinear modeling described Chapter 16 shall be used. The resulting force-displacement curve shall be used in lieu of
the assumed effective yield displacement, DY, of Equation 18.6-10 to calculate the effective ductility demand under the
design earthquake ground motion, µD, and under the risk-targeted maximum considered earthquake ground motion, µM,
in Equations 18.6-8 and 18.6-9, respectively. The value of (R/Cd) shall be taken as 1.0 in Equations 18.4-4, 18.4-5, 18.4-
8, and 18.4-9 for the response spectrum procedure, and in Equations 18.5-6, 18.5-7 and 18.5-15 for the equivalent lateral
force procedure.
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FIGURE 19.2-1, FOUNDATION DAMPING FACTOR
Chapter 19, Soil Structure Interaction for Seismic Design
TABLE 19.2-1, VALUES OF G / G0 AND Vs / Vso
Replace with the following:
Site Class
Value of vs / vs0
Value of G / G0
SDS/2.5
= 0.1
0.4
= 0.8
= 0.1
0.4
= 0.8
A
1.00
1.00
1.00
1.00
1.00
1.00
B
1.00
0.97
0.95
1.00
0.95
0.90
C
0.97
0.87
0.77
0.95
0.75
0.60
D
0.95
0.71
0.32
0.90
0.50
0.10
E
0.77
0.22
*
0.60
0.05
*
F
*
*
*
*
*
*
Note: Use straight line interpolation for intermediate values of SDS/2.5.
* Should be evaluated from site-specific analysis.
FIGURE 19.2-1, FOUNDATION DAMPING FACTOR
Replace with the following:
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Modification to Chapter 21, Site-Specific Ground Motion
Procedures for Seismic Design
SECTION 21.2, GROUND MOTION HAZARD ANALYSIS
Replace Sections 21.2.1 through 21.2.3 with the following:
21.2.1 Probabilistic Ground Motions. The probabilistic spectral response acceleration shall be taken as the
spectral response acceleration in the maximum direction of ground motions represented by a 5 percent damped
acceleration response spectrum that is expected to achieve a 1 percent probability of collapse within a 50-year
period. For the purpose of this provision, ordinates of the probabilistic ground-motion response spectrum shall
be determined by either Method 1 of Section 21.2.1.1 or Method 2 of Section 21.2.1.2.
21.2.1.1 Method 1. Ordinates of the probabilistic ground-motion response spectrum shall be determined as the
product of the risk coefficient at each spectral response period, CR, and the spectral response acceleration
represented by a 5 percent damped acceleration response spectrum having a 2 percent probability of exceedance
within a 50-year period. The value of the risk coefficient, CR, shall be determined using values of CRS and CR1
from Figures 22-3 and 22-4, respectively. At spectral response periods less than or equal to 0.2 second, CR shall
be taken as equal to CRS. At spectral response periods greater than or equal to 1.0 second, CR shall be taken as
equal to CR1. At response spectral periods greater than 0.2 second and less than 1.0 second, CR shall be based
on linear interpolation of CRS and CR1.
21.2.1.2 Method 2. Ordinates of the probabilistic ground-motion response spectrum shall be determined at
each spectral response period from the iterative integration of a site-specific hazard curve with a lognormal
probability density function representing the collapse fragility (i.e., probability of collapse as a function of
spectral response acceleration). At each period, the ordinate of the probabilistic ground-motion response
spectrum shall achieve a 1 percent probability of collapse within a 50-year period for a collapse fragility having
(i) a 10 percent probability of collapse at said ordinate of the probabilistic ground-motion response spectrum
and (ii) a logarithmic standard deviation value of 0.8.
21.2.2 Deterministic Ground Motions. The deterministic spectral response acceleration at each period shall
be calculated as the largest 84th percentile 5 percent damped spectral response acceleration in the direction of
maximum horizontal response computed at that period for characteristic earthquakes on all known active faults
within the region. For the purposes of this standard, the ordinates of the deterministic ground motions response
spectrum shall not be taken as lower than the corresponding ordinates of the response spectrum determined in
accordance with Figure 21.2-1, where Fa and Fv are determined using Tables 11.4-1 and 11.4-2, respectively,
with the value of Ss taken as 1.5 and the value of S1 taken as 0.6.
21.2.3 Site-Specific MCER. The site-specific MCER spectral response acceleration at any period, SaM, shall be
taken as the lesser of the spectral response accelerations from the probabilistic ground motions of Section 21.2.1
and the deterministic ground motions of Section 21.2.2.
Commentary to the Chapter 21 Modification
C21.2 GROUND MOTION HAZARD ANALYSIS
As explained in the commentary to Chapter 11, the risk-targeted maximum considered earthquake ground motions (MCER) in
the 2009 NEHRP Recommended Seismic Provisions are based on the 2008 USGS seismic hazard maps and also incorporate
three technical changes to ASCE/SEI 7-05:
1. Use of risk-targeted ground motions,
2. Use of maximum direction ground motions, and
3. Use of near-source 84th percentile ground motions.
Reasons for use of maximum direction ground motions are explained first in the commentary below, because they apply to
both the probabilistic and deterministic ground motions discussed subsequently. Use of risk-targeted and near-source 84th
percentile ground motions are discussed in the probabilistic and deterministic ground motions sections below, respectively.
The requirements in the previous editions of the Provisions and ASCE/SEI 7 do not define the direction of ground motions
used for design. The procedure used to develop the statistical estimate of ground motion results in the geometric mean
(geomean) of two orthogonal components of motion at a site. Many engineers find the maximum direction to be a more
meaningful parameter for structural design. The basic concept is that a structure is designed to resist the ground motion at its
site; the prediction of ground motion is inherently statistical, and the basis for the statistical estimate of the ground motion is
rooted in the probability that a structure will actually fail. In general, structures will not have the same resistance in all
directions; however, for those structures in which seismic resistance is a significant economic factor, there is a tendency to
design to the level required by building codes, with the result that the resistance of the structure is relatively insensitive to the
direction of the motion. When one considers such structures subjected to two simultaneous components of ground motion,
these structures characteristically fail in the direction of the stronger of the two components. Failure rates of simple buildings
in one recent study (low-rise wood buildings in Applied Technology Council, 2008) show this effect: the overall failure rate
for three-dimensional analyses was higher than those for two-dimensional analyses for the same set of structures analyzed for
the same 22 pairs of ground motions. The specification of maximum direction ground motions reduces the probability of
structural failure based upon equivalent static two-dimensional design compared to the use of the geomean based demand,
but this reduction has not been quantified at this time. For consistency, revisions have been made to both probabilistic and
deterministic ground motion criteria to reflect required use of maximum direction ground motions.
The USGS updates of the uniform-hazard and deterministic ground motion spectral value maps have used the new next
generation attenuation (NGA) relations for sites in the western United States (WUS). The new NGA relationships output an
average horizontal spectral demand and the dispersion in that demand, where this average is the rotated geomean denoted as
GMRotI50 (Boore et al., 2006). GM denotes the geometric mean of two horizontal components, Rot denotes that rotations
over all non-redundant angles are considered, I denotes that period-independent rotations are used, and 50 identifies the
prediction of median values. The geometric mean of two horizontal components of ground motions is calculated as the
square root of the product of the two horizontal response spectral accelerations at each period of interest. As demonstrated by
Boore et al. (2006), GMRotI50 is numerically very similar to (i.e., within 3 percent of) the geometric mean of two asrecorded
components that was typically the output of older attenuation relationships.
A recent study (Huang et al., 2008a) found that near-source ground motion spectral response accelerations of the new NGA
relations are somewhat less than those in the maximum direction of response. This study (2008a, 2008b) focused on large
magnitude earthquakes, with moment magnitudes greater than 6.5 and site-to-source distances less than 15 km. For this
family of earthquake records, ground motions in the maximum direction of response are about 110 percent of 5 percent
damped, short-period spectral response acceleration, and about 130 percent of 5 percent damped, 1-second spectral response
acceleration calculated using the new NGA relations (GMRotI50). Table C21.2-1 presents summary results to enable
calculation of median and 84th percentile ratios of maximum to geomean spectral demands across the period range of 0 to 4.0
seconds; values of the ratio are assumed to remain constant for periods greater than 4.0 seconds. Values are rounded to the
nearest 0.1, which is the appropriate degree of precision. The ratio of 84th percentile (Column 3) to median (Column 2)
demands is approximately 1.8 to 1.9. Linear interpolation should be used to establish values of the ratios for periods not
listed.
Other regions (e.g., the central and eastern United States) are expected to have similar ratios of maximum direction ground
motions to geomean ground motions although the limited number of strong-motion records from the central and eastern
United States precludes rigorous evaluation such as that performed by the NGA study (Huang et al., 2008). However, studies
by Beyer and Bommer (2006) using a set of 949 earthquake records with much wider ranges of moment magnitude (4.2 to
7.9) and hypocentral distance (5 to 200 km) indicated similar ratios of maximum to geomean response to those of the Huang
et al. study on large magnitude, near-fault ground motions. The Beyer and Bommer data set included records from 20+
European earthquakes.
Table C21.2-1 Median and 84th Percentile of the Ratio of
Maximum Spectral Demand to Geomean Demand
Period
(second)
Median
84th Percentile
Period
(second)
Median
84th Percentile
0.0
1.1
2.0
0.5
1.2
2.1
0.1
1.1
2.0
1.0
1.3
2.3
0.2
1.1
2.0
2.0
1.3
2.5
0.3
1.1
2.0
4.0+
1.4
2.7
For consistency of ground motion scaling (against either geomean or maximum direction spectra) in three-dimensional
response history analysis of structures, the 2009 Provisions has adopted changes related to Section 16.1.3.2 of ASCE/SEI 7-
05 such that it enables the scaling of pairs of horizontal ground motion records matching maximum direction spectra (MCER
or design spectra of maximum direction of response) to be equivalent to that matching the corresponding geomean spectra.
Additional explanation of these changes is provided in Section C16.1.3.2.
C21.2.1 Probabilistic Ground Motions. The definition and basis of probabilistic ground motions in these new Provisions
has changed from that in ASCE/SEI 7-05, from a 2 percent in 50-year hazard level to a 1 percent in 50-year collapse risk
target. This change is intended to improve seismic design by achieving a more uniform level of collapse prevention. The
change affects the calculation and values of probabilistic ground motions, but not their use in the design process (i.e., 5
percent damped spectral response accelerations are still used). The technical basis of the change can be found in “Risk-
Targeted versus Current Seismic Design Maps for the Conterminous United States” (Luco et al., 2007). A summary of the
technical basis is provided below.
In the 1997, 2000 and 2003 editions of the NEHRP Recommended Provisions, the probabilistic MCE ground motions are
defined as those that have a 2 percent probability of being exceeded in 50 years. In other words, the probabilistic MCE
ground motions are of uniform hazard, both geographically and across structural vibration periods. It has long been
recognized, however, that “it really is the probability of structural failure with resultant casualties that is of concern, and the
geographical distribution of that probability is not necessarily the same as the distribution of the probability of exceeding
some ground motion” (p. 296 of ATC 3-06, 1978).
The primary reason that the two probabilities are not the same is that there are geographic differences in the shape of the
ground motion versus annual frequency of exceedance hazard curves from which uniform-hazard ground motions are read.
The commentary of earlier editions of the Provisions (post-1997) reports that “because of these differences, questions were
raised concerning whether definition of the ground motion based on a constant probability for the entire United States would
result in similar levels of seismic safety for all structures” (p. 319 of the 2003 NEHRP Recommended Provisions
Commentary). The change to risk-targeted ground motions uses the different shapes of hazard curves to adjust the uniformhazard
(2-percent-in-50-years) ground motions such that they are expected to result in a uniform annual frequency of
collapse, or risk level, when used in design. The adjustment factors, or risk coefficients, are akin to the ASCE/SEI 43-05
site-specific design factor, which is a function of an approximate slope of the ground motion hazard curve.
The adjustments to the uniform-hazard ground motions are computed by making use of the so-called risk integral (e.g.,
McGuire, 2004). The risk integral calculates an annual frequency of collapse by coupling the ground motion hazard curve at
a location with the expected performance of a structure designed for that location. More precisely, the hazard curves are
coupled with the conditional probability of collapse as a function of the ground motion level. Earlier editions of the
Provisions express the expectation that “if a structure experiences a level of ground motion 1.5 times the design level [i.e.,
the MCE ground motion], the structure should have a low likelihood of collapse” (p. 320 of the 2003 NEHRP Provisions
Commentary). This “low likelihood of collapse” has been estimated as 10 percent (Applied Technology Council, 2009)
using state-of-the-art incremental dynamic analysis (e.g., Vamvatsikos and Cornell, 2002) of structures designed in
accordance with this edition of the NEHRP Recommended Seismic Provisions (2009). For the likelihood of collapse under
other (than the MCE) ground motion levels, a so-called ß-value of 0.8 has been used for the 2009 Provisions, based on both
the findings of the Applied Technology Council (2009) and other past research. Other ß-values ranging from 0.5 to 1.0 have
been considered, with little effect on the resulting risk coefficients. The ground motion hazard curves used in the risk integral
are from the USGS.
Using more subjective estimates of the conditional probability of collapse as a function of the ground motion level, and early
(1976) hazard curves for only four locations, the authors of the resource document on which the Provisions are based
(Applied Technology Council, 1978) used the risk integral to estimate the annual frequency of collapse of buildings designed
for uniform-hazard ground motions (see ATC 3-06, p. 310-311). They found that “the probabilities of failure [i.e., risk
levels] were roughly the same for each of the four buildings.” In contrast, using contemporary hazard curves and building
performance expectations, Luco et al. (2007) have found that the risk levels are systematically lower in the central and
eastern United States (CEUS) than in the WUS due to well-documented differences in the shapes of ground motion hazard
curves (e.g., Leyendecker et al., 2000). To result in uniform risk levels, adjustments to the uniform-hazard ground motions
are needed.
The risk level targeted in these Provisions (2009) corresponds (approximately) to 1 percent probability of collapse in 50
years. This target is based on the average of the annual frequencies of collapse across the WUS that are expected to result
from (as calculated via the risk integral) design for the probabilistic MCE ground motions in the 2003 NEHRP Recommended
Provisions. Consequently, the requisite risk coefficients are generally within 15 percent of unity in the WUS (except in the
coastal region of Oregon, where they are slightly smaller). In the CEUS, the risk coefficients are generally smaller, again due
to the well-documented differences in shapes of ground motion hazard curves there relative to the WUS. In the New Madrid
seismic zone and near Charleston, South Carolina, in particular, the adjustments to the uniform-hazard ground motions are as
small as a factor of 0.7. Compared to the underlying uniform-hazard ground motions, the risk coefficients are generally less
sensitive to refinements of the ground motion hazard curves (e.g., USGS updates or site-specific analyses), since they depend
on the shape but not amplitude of the hazard curves. They vary with the structural vibration period and site class, but not
dramatically.
The change to risk-targeted probabilistic ground motions complements improvements to the basis for response modification
factors (R factors) reflected in FEMA P-695 (Applied Technology Council, 2009) and provides a more rational basis for
seismic design methods. As alluded to above, similar risk-based procedures are already being used for design and evaluation
of nuclear facilities, as well as offshore structures.
C21.2.2 Deterministic Ground Motions. Deterministic ground motions should account for uncertainties associated with
near-fault ground motions, particularly at longer periods, and necessitate a more statistically appropriate estimate of 5 percent
damped spectral response accelerations than those based on the 150 percent of the median ground motions used in ASCE/SEI
7-05. The use of 84th percentile ground motions in these Provisions (2009) effectively requires increasing median ground
motions by 180 percent. The technical basis of this change can be found in Huang et al. (2008a and 2008b). The authors
found that 150 percent of the median spectral response accelerations of the new NGA relations (average of the three
relations) to be significantly less than 84th percentile ground motions in the maximum direction of response. Near active
sources (in the WUS), 84th percentile ground motion in the maximum direction of response is about 200 percent (1.8 x 110
percent) of 5 percent damped, short-period spectral response acceleration, and about 230 percent (1.8 x 130 percent) of 5
percent damped, 1-second spectral response acceleration of the new NGA relations for GMRotI50 (average value of the three
NGA relations). Table C21.2-2 summarizes ratios of 84th percentile maximum direction to median geomean-direction
response for periods from 0 to 4.0 seconds. Ratios for periods greater than 4.0 seconds are assumed to be the same as the
ratio for 4.0 seconds.
Table C21.2-2 Ratios of 84th Percentile to Median Spectral Demands for NGA Relationships
Period (seconds)
0.2
0.5
1.0
2.0
3.0
4.0
Equation
ß
B-A
0.60
0.61
0.65
0.70
0.70
0.70
C-B
0.59
0.59
0.62
0.64
0.65
0.65
C-Y
0.61
0.63
0.63
0.67
0.67
0.70
Equation
y84 / y50
1.82
1.84
1.89
1.95
1.96
1.98
ADDITIONAL REFERENCES FOR CHAPTER 21 COMMENTARY
Abrahamson, N., and W. J. Silva. 1997. "Empirical Response Spectral Attenuation Relations for Shallow Crustal
Earthquakes." Seismological Research Letters, 68(1):94-127.
American Society of Civil Engineers. 2006. Minimum Design Loads for Buildings and Other Structures, ASCE/SEI 7-05.
ASCE, Reston, Virginia.
American Society of Civil Engineers. 2005. Seismic Design Criteria for Structures, Systems, and Components in Nuclear
Facilities, ASCE 43-05. ASCE, Reston, Virginia.
Applied Technology Council. 1978. Tentative Provisions for the Development of Seismic Regulations for Buildings, ATC 3-
06. ATC, Palo Alto, California.
Applied Technology Council. 2009. Quantification of Building Seismic Performance Factors, FEMA P-695. Federal
Emergency Management Agency, Washington, D.C.
Beyer, K., and J. J. Bommer. 2006. Relationships between Median Values and between Aleatory Variabilities for Different
Definitions of the Horizontal Component of Motion, Bulletin of the Seismological Society of America, 96(4A):1512-1522.
Boore, D. M., and T. E. Fumal. 1997. "Equations for Estimating Horizontal Response Spectra and Peak Acceleration from
Western North American Earthquakes: A Summary of Recent Work," Seismological Research Letters, 68(1):128-153.
Boore, D. M., J. Watson-Lamprey, and N. A. Abrahamson. 2006. "Orientation-Independent Measures of Ground Motion,"
Bulletin of the Seismological Society of America, 96(4A):1502-1511.
Boore, D. M., and G. M. Atkinson. 2007. Boore-Atkinson NGA Ground Motion Relations for the Geometric Mean
Horizontal Component of Peak And Spectral Ground Motion Parameters, PEER 2007/01. Pacific Earthquake Engineering
Research Center, Berkeley, California.
Campbell, K. W., and Y. Bozorgnia. 2003. "Updated Near-Source Ground Motion (Attenuation) Relations for the
Horizontal and Vertical Components of Peak Ground Acceleration and Acceleration Response Spectra." Bulletin of the
Seismological Society of America, 93:314-331.
Campbell, K. W., and Y. Bozorgnia. 2007. Campbell-Bozorgnia NGA Ground Motion Relations for the Geometric Mean
Horizontal Component of Peak and Spectral Ground Motion Parameters," PEER 2007/02. Pacific Earthquake Engineering
Research Center, Berkeley, California.
Chiou, B. S.-J., and R. R. Youngs. 2006. Chiou and Youngs PEER-NGA Empirical Ground Motion Model for the Average
Horizontal Component of Peak Acceleration and Pseudo-Spectral Acceleration for Spectral Periods of 0.01 to 10 Seconds.
Pacific Earthquake Engineering Research Center, Berkeley, California, http://peer.berkeley.edu/products/CYProgram/
Chiou_Youngs_NGA_2006.pdf.
Huang, Y.-N, A. S. Whittaker, and N. Luco. 2008a. NGA Relationships, USGS Seismic Hazard Maps, Near-Fault Ground
Motions and Site Effects, USGS, Golden, Colorado.
Huang, Y.-N, A. S. Whittaker, and N. Luco. 2008b. “Maximum Spectral Demands in the Near-Fault Region,” Earthquake
Spectra, 24(1):319-341.
Leyendecker, E.V., R. J. Hunt, A. D. Frankel, and K. S. Rukstales. 2000. “Development of Maximum Considered
Earthquake Ground Motion Maps,” Earthquake Spectra, 16(1):21-40.
Luco, N., B. R. Ellingwood, R. O. Hamburger, J. D. Hooper, J. K. Kimball, and C. A. Kircher. 2007. “Risk-Targeted versus
Current Seismic Design Maps for the Conterminous United States,” in Proceedings of the SEAOC 76th Annual Convention.
Structural Engineers Association of California, Sacramento, California.
McGuire, R. K. 2004. Seismic Hazard and Risk Analysis,” EERI Monograph MNO-10. Earthquake Engineering Research
Institute, Oakland, California.
Sadigh, K., C. Y. Chang, J. A. Egan, F. Makdisi, and R. R. Youngs. 1997. "Attenuation Relationships for Shallow Crustal
Earthquakes Based on California Strong Motion Data." Seismological Research Letters, 68(1):180-189.
Somerville, P. G., N. F. Smith, R. W. Graves, and N. A. Abrahamson. 1997. "Modification of Empirical Strong Ground
Motion Attenuation Relations To Include the Amplitude And Duration Effects of Rupture Directivity," Seismological
Research Letters, 68(1):199-222.
Vamvatsikos, D., and C. A. Cornell. 2002. “Incremental Dynamic Analysis,” in Earthquake Engineering and Structural
Dynamics, 31(3):491-514.
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Modification to Chapter 22, Seismic Ground Motion and
Long-period Transition Maps
Replace existing Chapter 22 with the following:
Chapter 22
SEISMIC GROUND MOTION, LONG-PERIOD TRANSITION,
RISK COEFFICIENT, AND MCE GEOMEAN PGA MAPS
Contained in this chapter are Figures 22-1 through 22-7, which provide the mapped uniform-hazard ground motion
parameters (SSUH and S1UH), the mapped risk coefficients (CRS and CR1), the mapped deterministic ground motion
parameters (SSD and S1D), and the mapped long-period transition period (TL), for use in applying the seismic
provisions of ASCE/SEI 7. Also contained in this chapter are Figures 22-8 through 22-11, which provide the
mapped maximum considered earthquake geometric mean peak ground accelerations.
These maps were prepared by the United States Geological Survey (USGS) and have been updated for the 2009
edition of the NEHRP Recommended Seismic Provisions for New Buildings and Other Structures. Maps for Guam
and Tutuila (American Samoa) are not included because uniform-hazard ground motion parameters, deterministic
ground motion parameters, and risk coefficients have not yet been developed for those islands. Therefore, like in the
2005 edition of ASCE/SEI 7, the parameters SS and S1 defined in Section 11.4.3 shall be, respectively, 1.5 and 0.6
for Guam and 1.0 and 0.4 for Tutuila. The mapped maximum considered earthquake geometric mean peak ground
accelerations shall be 0.6 for Guam and 0.4 for Tutuila.
The following is a list of figures contained in this chapter:
Figure 22-1 Uniform-hazard (2% in 50-year) ground motions of 0.2-second spectral response acceleration (5%
of critical damping), Site Class B.
Figure 22-2 Uniform-hazard (2% in 50-year) ground motions of 1-second spectral response acceleration (5%
of critical damping), Site Class B.
Figure 22-3 Risk coefficient at 0.2-second spectral response period.
Figure 22-4 Risk coefficient at 1-second spectral response period.
Figure 22-5 Deterministic ground motions of 0.2-second spectral response acceleration (5% of critical
damping), Site Class B.
Figure 22-6 Deterministic ground motions of 1-second spectral response acceleration (5% of critical damping),
Site Class B.
Figure 22-7 Long-period transition period, TL (seconds).
Figure 22-8 MCE geometric mean PGA, %g, Site Class B for the coterminous United States.
Figure 22-9 MCE geometric mean PGA, %g, Site Class B for Alaska.
Figure 22-10 MCE geometric mean PGA, %g, Site Class B for Hawaii.
Figure 22-11 MCE geometric mean PGA, %g, Site Class B for Puerto Rico and the United States Virgin
Islands.
Figure22-1 Map of the US
Figure22-1 (cont) Map of the US
Figure 22-2 Map of the US
Figure 22-2 (cont) Map of the US
Figure 22-3 Map of the US
Figure 22-3 (cont) Map of the US
Figure 22-4 Map of the US
Figure 22-4 (cont) Map of the US
Figure 22-5 Map of the US
Figure 22-5 (cont) Map of the US
Figure 22-6 Map of the US
Figure 22-6 (cont) Map of the US
Figure 22-7 Map of the US
Figure 22-7 (cont) Map of the US
Figure 22-8 Map of the US
Figure 22-8 (cont) Map of the US
Figures 22-9 and 10 Map of the US
Figures 22-11 showing a US Map
Commentary to New Chapter 22
Chapter 22 Commentary
SEISMIC GROUND MOTION, LONG-PERIOD TRANSITION,
RISK COEFFICIENT, AND MCE GEOMEAN PGA MAPS
The USGS has prepared the four new sets of maps for Chapter 22 of the 2009 NEHRP Recommended Seismic Provisions:
1. Maps of uniform-hazard (2 percent in 50-year) ground motions,
2. Maps of the risk coefficients for converting 2 percent in 50-year uniform-hazard ground motions to 1 percent in 50-year
risk-targeted probabilistic ground motions,
3. Maps of deterministic ground motions (consistent with site-specific criteria of Section 21.2.2), and maps of peak ground
accelerations for the evaluation of the potential for liquefaction and soil strength loss (according to Section 11.8.3).
Because this would have resulted in a substantial increase in the number of maps, the BSSC Provisions Update Committee
recommended that the separate maps for regions of the United States and its territories that appeared in ASCE/SEI 7-05 be
consolidated (for the uniform-hazard ground motion, risk coefficient, and deterministic ground motion maps), into the single
figures in Chapter 22. Thus, the total number of map figures (11) in these Provisions (2009) is less than that in ASCE/SEI 7-
05 (i.e., 20). Because the consolidated map figures are relatively small and difficult to read, the USGS website that
automates use of the maps and formulas will be especially useful (http://earthquake.usgs.gov.designmaps/usapp).
As described in the commentary to Chapter 21 and below, the uniform-hazard and deterministic ground motion maps in
Chapter 22 of these Provisions (2009) represent response in the maximum direction. The USGS has developed these maps
based on "geomean" ground motions (the product of hazard assessment using modern ground motion attenuation functions),
adjusted using constant factors that transform geomean response to maximum direction response. The same factors (i.e., 1.1
at short-periods and 1.3 at a period of 1 second) are used for all seismic regions (i.e., both the central and eastern United
States or CEUS and the western United States or WUS) and for both probabilistic and deterministic ground motions.
In contrast, the peak ground acceleration maps in Chapter 22 represent geomean ground motions, as described below.
Furthermore, the peak ground acceleration maps represent the lesser of uniform-hazard (2 percent in 50-year) and
deterministic peak ground accelerations, without consideration of corresponding risk coefficients.
Uniform-Hazard (2 Percent in 50-Year) Ground Motion Maps
The uniform-hazard maps in Chapter 22 of these Provisions (2009) are based on the 2008 USGS National Seismic Hazard
Maps (http://earthquake.usgs.gov/hazmaps); however, since the ground motion values on the uniform-hazard maps are for the
maximum direction of acceleration (as explained above), they are different from the “geomean” USGS maps. The 0.2-
second and 1-second spectral response acceleration uniform-hazard maps are different by a factor of 1.1 and 1.3 from the
respective USGS maps. Development of the USGS maps is documented in Petersen et al. (2008).
Risk Coefficient Maps
Development of risk coefficients and related work by the USGS is documented by Luco et al. (2007). The risk coefficient
maps indicate that, in general, risk-targeted probabilistic ground motions (based on 1 percent in 50-year collapse risk) would
moderately decrease the uniform-hazard ground motions (based on 2 percent in 50-year hazard) in high-hazard areas of the
CEUS and the coastal region of Oregon (by as much as 30 percent) and either slightly increase or decrease the uniformhazard
ground motions in the WUS and remaining areas of the maps (by less than 15 percent). These changes do not affect
calculation of deterministic ground motions, which often govern in high seismic areas.
Deterministic Ground Motion Maps
The deterministic maps in Chapter 22 of the Provisions represent the greater of 84th percentile (maximum direction)
response and the “water level” values described in the next paragraph. The USGS has developed these maps based on
median "geomean" ground motions (the product of hazard assessment using modern ground motion attenuation functions)
adjusted using factors that transform median geomean-direction response to 84th percentile maximum-direction response.
The same factors (i.e., 1.1 x 1.8 at short-periods and 1.3 x 1.8 at a period of 1 second) are used for all seismic regions (i.e.,
both the CEUS and WUS regions).
As defined in ASCE/SEI 7-05 Section 21.2.2, the deterministic spectral response accelerations (for Site Class B) shall not be
taken as lower than 1.5g for the short periods and 0.6g for the 1-second period; hence, the ground motions on the
deterministic maps (Figures 22-3 and 22-4) are no lower than these values. Otherwise the ground motions on the
deterministic maps are 180 percent (as opposed 150 percent in ASCE/SEI 7-05) of median spectral response accelerations,
for reasons explained above in the commentary to Chapter 21. Like the uniform-hazard maps described above, the
deterministic maps represent the spectral response acceleration in the maximum direction.
Peak Ground Acceleration Maps
Unlike the uniform-hazard and deterministic ground motion maps described above, the peak ground acceleration maps in
Chapter 22 of the Provisions represent geometric mean ground motions (not response in the maximum direction). Despite
representing geometric mean ground motions, the peak ground acceleration maps are different from the 2008 USGS National
Seismic Hazard Maps (http://earthquake.usgs.gov/hazmaps) upon which they are based. This is because they represent the
lesser of uniform-hazard (2 percent in 50-year hazard) and deterministic peak ground accelerations. Development of the
uniform-hazard peak ground accelerations is documented in Petersen et al. (2008). The deterministic peak ground
accelerations are calculated as the greater of 180 percent of median ground motions and a water level of 0.6g, Note that risk
coefficients are not included in the development of the peak ground acceleration maps, which is why they are referred to as
“maximum considered earthquake geometric mean peak ground acceleration” maps without the “risk-targeted” prefix.
REFERENCES
American Society of Civil Engineers. 2006. Minimum Design Loads for Buildings and Other Structures, ASCE/SEI 7-05.
ASCE, Reston, Virginia.
Luco, N., B. R. Ellingwood, R. O. Hamburger, J. D. Hooper, J. K. Kimball and C. A. Kircher. 2007. “Risk-Targeted versus
Current Seismic Design Maps for the Conterminous United States,” in Proceedings of the SEAOC 76th Annual Convention.
Structural Engineers Association of California, Sacramento, California.
Petersen, M. D., A. D. Frankel, S. C. Harmsen, C. S. Mueller, K. M. Haller, R. L. Wheeler, R. L. Wesson, Y. Zeng, O. S.
Boyd, D. M. Perkins, N. Luco, E. H. Field, C. J. Wills, and K. S. Rukstales. 2008. Documentation for the 2008 Update of
the United States National Seismic Hazard Maps, USGS Open File Report 2008-1128. USGS, Golden, Colorado.
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Modifications to Chapter 23,
Seismic Design Reference Documents
SECTION 23.1, CONSENSUS STANDARDS AND OTHER REFERENCE DOCUMENTS
Add the following entries:
ASCE 41
Supplement 1, Section 3.3.3
Seismic Rehabilitation of Existing Buildings, 2007
ANSI/AISI S110
Sections 14.1.1, 14.1.2, 14.1.3, Table 12.2-1
Standard for Seismic Design of Cold-Formed Steel Structural Systems – Special Bolted Moment Frames, 2007.
ANSI/RMI MH 16.1
Section 15.5.3
Specification for the Design, Testing, and Utilization of Industrial Steel Storage Racks, 2008
Revise the following entries to read as indicated:
ACI 318
Sections 14.2.2, 14.2.2.1, 14.2.2.2, 14.2.2.3, 14.2.2.4, 14.2.2.5, 14.2.2.6, 14.2.2.7, 14.2.2.8, 14.2.2.9, 14.2.3, 14.2.3.1.1,
14.2.3.2.1, 14.2.3.2.2, 14.2.3.2.3, 14.2.3.2.5, 14.2.3.2.
Building Code Requirements for Structural Concrete, 2008.
NFPA 13
Sections 13.6.5.1, 13.6.8, 13.6.8.2, 13.6.8.4
Standard for the Installation of Sprinkler Systems, 2007
Delete the following entry:
RMI
Rack Manufacturers Institute
8720 Red Oak Boulevard
Suite 201
Charlotte, NC 28217
RMI
Section 15.5.3
Specification for the Design, Testing, and Utilization of Industrial
Steel Storage Racks 1997, reaffirmed 2002
Page intentionally left blank.
New Chapter 23, Vertical Ground Motions for Seismic Design
Add the following new Chapter 23 and renumber the existing ASCE/SEI 7-05 Chapter 23 as
Chapter 24:
Chapter 23
VERTICAL GROUND MOTIONS FOR SEISMIC DESIGN
23.1 DESIGN VERTICAL RESPONSE SPECTRUM. Where a design vertical response spectrum is required by
these Provisions and site-specific procedures are not used, the design vertical response spectral acceleration, Sav, (in g –
gravity unit) shall be developed as follows:
1. For vertical periods less than or equal to 0.025 second, Sav shall be determined in accordance with Equation 23.1-1
as follows:
Sav = 0.3CVSDS (23.1-1)
2. For vertical periods greater than 0.025 second and less than or equal to 0.05 second, Sav shall be determined in
accordance with Equation 23.1-2 as follows:
Sav = 20CVSDS(TV - 0.025)+0.3CVSDS (23.1-2)
3. For vertical periods greater than 0.05 second and less than or equal to 0.15 second, Sav shall be determined in
accordance with Equation 23.1-3 as follows:
Sav = 0.8CVSDS (23.1-3)
4. For vertical periods greater than 0.15 second and less than or equal to 2.0 seconds, Sav shall be determined in
accordance with Equation 23.1-4 as follows:
(23.1-4)
where CV is defined in terms of SS in Table 23.1-1, SDS = the design spectral response acceleration parameter at short
periods, and TV = the vertical period of vibration.
Table 23.1-1 Values of Vertical Coefficient CV Equation
0.75 0.8 0.15 av V DS
V
S C S
T
. .
= . .
. .
MCER Spectral
Response Parameter at
Short Periods a
Site Class A, B
Site Class C
Site Class D, E, F
Ss = 2.0
0.9
1.3
1.5
SS = 1.0
0.9
1.1
1.3
SS = 0.6
0.9
1.0
1.1
SS = 0.3
0.8
0.8
0.9
SS = 0.2
0.7
0.7
0.7
a Use straight-line interpolation for intermediate values of SS
.
Sav shall not be less than one-half (1/2) of the corresponding Sa for horizontal components determined in accordance with
the general or site-specific procedures of Section 11.4 or Chapter 21, respectively.
For vertical periods greater than 2.0 seconds, Sav shall be developed from a site-specific procedure; however, the
resulting ordinate of Sav shall not be less than one-half (1/2) of the corresponding Sa for horizontal components
determined in accordance with the general or site-specific procedures of Section 11.4 or Chapter 21, respectively.
In lieu of using the above procedure, a site-specific study may be performed to obtain Sav at vertical periods less than or
equal to 2.0 seconds, but the value so determined shall not be less than 80 percent of the Sav value determined from
Equations 23.1-1 through 23.1-4.
23.2 MCER VERTICAL RESPONSE SPECTRUM. The MCER vertical response spectral acceleration shall be 150
percent of the Sav determined in Section 23.1.
Commentary to New Chapter 23
Chapter 23 Commentary
VERTICAL GROUND MOTIONS FOR SEISMIC DESIGN
C23.1 DESIGN VERTICAL RESPONSE SPECTRUM
General. ASCE/SEI 7-05 and the earlier editions of the Provisions use the term 0.2SDSD to reflect the effects of vertical
ground motion. Where a more explicit consideration of vertical ground motion effects is advised—as for certain tanks,
materials storage facilities, and electric power generation facilities—the requirements of this chapter may be applied.
Historically, the amplitude of vertical ground motion has been inferred to be two-thirds (2/3) the amplitude of the horizontal
ground motion. However, studies of horizontal and vertical ground motions over the past 25 years have shown that such a
simple approach is not valid in many situations (e.g., Bozorgnia and Campbell, 2004, and references therein) for the
following main reasons: (a) vertical ground motion has a larger proportion of short-period (high-frequency) spectral content
than horizontal ground motion and this difference increases with decreasing soil stiffness and (b) vertical ground motion
attenuates at a higher rate than horizontal ground motion and this difference increases with decreasing distance from the
earthquake.
The observed differences in the spectral content and attenuation rate of vertical and horizontal ground motion lead to the
following observations regarding the vertical/horizontal (V/H) spectral ratio (Bozorgnia and Campbell, 2004):
1. The V/H spectral ratio is relatively sensitive to spectral period, distance from the earthquake, local site conditions, and
earthquake magnitude (but only for relatively soft sites) and relatively insensitive to earthquake mechanism and sediment
depth;
2. The V/H spectral ratio has a distinct peak at short periods that generally exceeds 2/3 in the near-source region of an
earthquake; and
3. The V/H spectral ratio is generally less than 2/3 at mid-to-long periods.
Therefore, depending on the period, the distance to the fault, and the local site conditions of interest, use of the traditional
2/3V/H spectral ratio can result in either an underestimation or an overestimation of the expected vertical ground motions.
The procedure for defining the design vertical response spectrum in the Provisions is based on the studies of horizontal and
vertical ground motions conducted by Campbell and Bozorgnia (2003) and Bozorgnia and Campbell (2004). These
procedures are also generally compatible with the general observations of Abrahamson and Silva (1997) and Silva (1997) and
the proposed design procedures of Elnashai (1997).
General Design Procedure. In order to be consistent with the shape of the horizontal design response spectrum, the vertical
design response spectrum has four regions defined by the vertical period of vibration (Tv). Based on the study of Bozorgnia
and Campbell (2004), the periods that define these regions are approximately constant with respect to the magnitude of the
earthquake, the distance from the earthquake, and the local site conditions. In this respect, the shape of the vertical response
spectrum is simpler than that of the horizontal response spectrum.
The equations that are used to define the design vertical response spectrum are based on three observations made by
Bozorgnia and Campbell (2004):
1. The short-period part of the 5 percent damped vertical response spectrum is controlled by the spectral acceleration at Tv =
0.1 second;
2. The mid-period part of the vertical response spectrum is controlled by a spectral acceleration that decays as the inverse
of the 0.75 power of the vertical period of vibration (Tv
-0.75); and
3. The short-period part of the V/H spectral ratio is a function of the local site conditions, the distance from the earthquake
(for sites located within about 60 km of the fault), and the earthquake magnitude (for soft sites).
The Provisions do not include seismic design maps for the vertical spectral acceleration at Tv = 0.1 second and do not
preserve any information on the earthquake magnitudes or the source-to-site distances that contribute to the horizontal
spectral accelerations that are mapped. Therefore, the general procedure recommended by Bozorgnia and Campbell (2004)
was modified to use only those horizontal spectral accelerations that are available from the seismic design maps, as follows:
1. Estimate the vertical spectral acceleration at Tv = 0.1 second from the ratio of this spectral acceleration to the horizontal
spectral acceleration at T = 0.2 second for the Site Class BC boundary (i.e., the boundary between Site Classes B and C (
m/sec), the reference site condition for the 2008 U.S. Geological Survey National Seismic Hazard Maps). For
earthquakes and distances for which the vertical spectrum might be of engineering interest (magnitudes greater than 6.5
and distances less than 60 km), this ratio is approximately 0.8 for all site conditions (Campbell and Bozorgnia, 2003). Equation
760 s v =
2. Estimate the horizontal spectral acceleration at T = 0.2 second from the Next Generation Attenuation (NGA) relationship
of Campbell and Bozorgnia (2008) for magnitudes greater than 6.5 and distances ranging between 1 and 60 km for the
Site Class BC boundary ( m/sec). The relationship of Campbell and Bozorgnia (2008), rather than that of
Campbell and Bozorgnia (2003), was used for this purpose in order to be consistent with the development of the 2008
U.S. Geological Survey National Seismic Hazard Maps, which use the NGA attenuation relationships to estimate
horizontal ground motions in the western United States. Similar results were found for the other two NGA relationships
that were used to develop the seismic hazard and design maps (Boore and Atkinson, 2008; Chiou and Youngs, 2008). Equation
760 s v =
3. Use the dependence between the horizontal spectral acceleration at T = 0.2 second and source-site distance estimated in
Item 2 and the relationship between the V/H spectral ratio, source-site distance, and local site conditions in Bozorgnia
and Campbell (2004) to derive a relationship between the vertical spectral acceleration and the mapped MCER spectral
response acceleration parameter at short periods, SS.
4. Use the dependence between the vertical spectral acceleration and the mapped MCER spectral response acceleration
parameter at short periods, SS, in Item 3 to derive a vertical coefficient, Cv, that when multiplied by 0.8 and the design
horizontal response acceleration at short periods, SDS, results in an estimate of the design vertical spectral acceleration at
Tv = 0.1 second.
Detailed Design Procedure. The following description of the detailed design procedure listed in Section 23.1 refers to the
illustrated design vertical response spectrum in Figure C23.1-1.
Vertical periods less than or equal to 0.025 second. Equation 23.1-1 defines that part of the design vertical response
spectrum that is controlled by the vertical peak ground acceleration. The 0.3 factor was approximated by dividing the 0.8
factor that represents the ratio between the vertical spectral acceleration at Tv = 0.1 second and the horizontal spectral
acceleration at T = 0.2 second by 2.5, the factor that represents the ratio between the design horizontal spectral acceleration at
T = 0.2 second, SDS, and the zero-period acceleration used in the development of the design horizontal response spectrum.
The vertical coefficient, Cv, in Table 23.1-1 accounts for the dependence of the vertical spectral acceleration on the amplitude
of the horizontal spectral acceleration and the site dependence of the V/H spectral ratio as determined in Items 3 and 4 above.
The factors are applied to SDS rather than to SS because SDS already includes the effects of local site conditions and the 2/3
factor that is required to reduce the horizontal spectral acceleration from its MCER value to its design value.
Vertical periods greater than 0.025 second and less than or equal to 0.05 second. Equation 23.1-2 defines that part of the
design vertical response spectrum that represents the linear transition from the part of the spectrum that is controlled by the
vertical peak ground acceleration and the part of the spectrum that is controlled by the dynamically amplified short-period
spectral plateau. The factor of 20 is the factor that is required to make this transition continuous and piecewise linear
between these two adjacent parts of the spectrum.
Vertical periods greater than 0.05 second and less than or equal to 0.15 second. Equation 23.1-3 defines that part of the
design vertical response spectrum that represents the dynamically amplified short-period spectral plateau.
Vertical periods greater than 0.15 second and less than or equal to 2.0 seconds. Equation 23.1-4 defines that part of the
design vertical response spectrum that decays with the inverse of the vertical period of vibration raised to the 0.75 power.
Limits Imposed on Sav. Two limits are imposed on the design vertical response spectrum defined by Equations 23.1-1
through 23.1-4 and illustrated in Figure 23.1-1. The first limit restricts the vertical period of vibration to be no larger than 2
seconds. This limit accounts for the fact that such large vertical periods are rare (structures are inherently stiff in the vertical
direction) and that the vertical spectrum might decay differently with period at longer periods. There is an allowance for
developing a site-specific design vertical response spectrum if this limit is exceeded (see Section 11.4 or Chapter 21 for
guidance on applying site-specific methods). The second limit restricts the design vertical response spectrum to be no less
than 50 percent of the design horizontal response spectrum. This limit accounts for the fact that a V/H spectral ratio of onehalf
(1/2) is a reasonable, but somewhat conservative, lower bound over the period range of interest, based on the results of
Campbell and Bozorgnia (2003) and Bozorgnia and Campbell (2004).
Figure C23.1-1 An Illustrative example of the design vertical response spectrum.
0.5 1.0 1.5 2.0
Vertical Period, Tv (sec)
Vertical Spectral Acceleration
0.15
0.05
0.3 CV SDS
0.8 CV SDS
0.8 CV SDS (0.15/TV )0.75
0.025
Figure C23.1-1 Illustrative example of the design vertical response spectrum.
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